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Behavior of preloaded geosynthetic-reinforced soil mass

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Title:
Behavior of preloaded geosynthetic-reinforced soil mass
Creator:
Ketchart, Kanop
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English
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347 leaves : ; 28 cm

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Subjects / Keywords:
Geosynthetics ( lcsh )
Soil stabilization ( lcsh )
Soils -- Creep ( lcsh )
Geosynthetics ( fast )
Soil stabilization ( fast )
Soils -- Creep ( fast )
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bibliography ( marcgt )
theses ( marcgt )
non-fiction ( marcgt )

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Bibliography:
Includes bibliographical references (leaves 335-347).
General Note:
Department of Civil Engineering
Statement of Responsibility:
by Kanop Ketchart.

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University of Colorado Denver
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Auraria Library
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All applicable rights reserved by the source institution and holding location.
Resource Identifier:
45211847 ( OCLC )
ocm45211847
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LD1190.E53 2000d .K48 ( lcc )

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Full Text
BEHAVIOR OF PRELOADED GEOSYNTHETIC-REINFORCED
SOIL MASS
by
Kanop Ketchart
B.Eng., Chulalongkom University, 1992
M.S., University of Colorado at Denver, 1995
A thesis submitted to the
University of Colorado at Denver
in partial fulfillment
of the requirements for the degree of
Doctor of Philosophy
Civil Engineering
2000


This thesis for the Doctor of Philosophy
degree by
Kanop Ketchart
has been approved
by

Jdb4than JT.H. Wu


Ketchart, Kanop (Ph.D., Civil Engineering)
Behavior of Preioaded Geosynthetic-Reinforced Soil Mass
Thesis directed by Professor Jonathan T.H. Wu
ABSTRACT
A study was undertaken to investigate the behavior of a geosynthetic-
reinforced soil (GRS) mass subject to preloading and to develop a simplified
analytical model for estimating deformation of a preloaded GRS mass. Since 1997,
preloading has been applied to a number of full-scale GRS structures. The preloading
has been shown to reduce setdements in the subsequent loading paths. However, the
behavior of a preloaded GRS mass has not yet been fully elucidated. Specifically,
two important questions remain unanswered:
1. What is the appropriate loading magnitude and loading sequence to effectively
preload a GRS mass?
2. How much deformation is to be expected in a preloaded GRS mass?
To seek answers to these questions, a systematic study was conducted. A
series of laboratory tests were performed on a number of soils, geosynthetics, and
soil-geosynthetic interfaces under different loading-unloading sequences. A modified
soil-geosynthetic performance (SGP) test apparatus was devised for investigating the
m


behavior of a generic GRS mass. The results of the SGP tests performed on various
soil-geosyntheic composites provide answers to the first question.
Some major findings from the experimental study are:
(1) Preloading significantly reduced vertical and horizontal deformations of the GRS
mass in subsequent loading paths by factors of 2 to 7.
(2) The reloading stiffness in the vertical direction increased by factors of 2 to 2.5
with increasing prestressed load level.
(3) The prestressed load level did not affect the horizontal reloading stiffness.
To answer the second question, a simplified preloading-reloading (SPR)
model was developed. The SPR model was developed based on the elastic analysis
of an idealized plane-strain GRS mass. The SPR model is shown to be capable of
predicting the experimental and numerical analysis results of the APSR Test and the
SGP test results conducted in this study. A parametric study was conducted with the
SPR model. The study showed that the vertical displacement of the GRS mass was
significantly reduced by increasing the soil stiffness. The effect of increasing the
reinforcement stiffness was much smaller.
Correlations between SGP test and preloaded GRS structures were proposed
and evaluated by two preloaded full-scale GRS structures. It was found that the
degree of reduction in the settlement due to preloading of GRS structures may be
assessed with reasonable accuracy by the results of a SGP test or simply by using the
SPR model.
IV


This abstract accurately represents the content of the candidates thesis. I recommend
its publication. ,
Signed

Jonathan T.H. Wu
v


ACKNOWLEDGMENT
I would like to express my sincere thanks and appreciation to my advisor
Professor Jonathan Til. Wu for his guidance, inspiration, support, and friendship. It
has been a most rewarding experience working with him. I would also like to thank
Michael T. Adams of the Tumer-Fairbank Highway Research Center, Federal
Highway Administration, for his friendship, encouragement, and insightful
discussions over the years. My thanks extend to professors Hon-Yim Ko, Stein Sture,
Dunja Peric', and John Trapp for serving on the examining committee. Financial
support from the Federal Highway Administration under a grant titled "Performance
Test for Reinforced Soils" is gratefully acknowledged.
Finally, my thanks go to my beloved parents and sister for their constant love
and support throughout my entire education.


CONTENTS
Figures......................................................................jd
Tables.....................................................................jcix
Chapter
1. Introduction..............................................................I
1.1 Problem Statement.........................................................I
12. Research Objectives......................................................3
13 Method of Research.........................................................3
2. Literature Review.........................................................5
2.1 Behavior of Sand subject to Unloading-Reloading Cycles-------------------5
2.2 Behavior of Geosynthetic subject to Unloading-Reloading Cycles..........13
2.3 Behavior of Soil-Geosynthetic Interface subject to
Unloading-Reloading Cycles___________________________________________________17
2.4 Behavior of GRS Mass subject to Unloading-Reloading Cycles--------------17
2.4.1 General Behavior.......................................................17
2.4.2 Preloaded GRS Structures...............................................24
2.5 Plane Strain Tests for Reinforced-Soii Mass.............................37
3. Laboratory Tests on Soils, Geosynthetics, and
Soil-Geosynthetic Interfaces_________________________________________________50
3.1 Test materials___________________________________________________________51
3.1.1 Soils__________________________________________________________________51
3.1.2 Geosynthetics__________________________________________________________52
3 2 Loading System__________________________________________________________56
33 Loading Sequences_________________________________________________________57
vii


3.4 Conventional Triaxial Compression Tests for Soils......................61
3.4.1 Test Description......................................................61
3.42 Specimen Preparation and Test Procedure...............................63
3.4.3 Measurement and Data Reduction.......................................65
3.4.4 Test Programs.........................................................67
3.4.5 Test Results and Discussions..........................................70
3.5 In-Isolation Load-Extension Tests for Geosynthetics------------------93
3.5.1 Test Description......................................................93
3.5.2 Specimen Preparation and Test Procedure...............................93
3.5.3 Measurement and Data Reduction........................................95
3.5.4 Test Programs........................................................101
3.5.5 Test Results and Discussions.........................................101
3.6 Interface Direct Shear Tests for Soil-Geosynthetic Interfaces---------115
3.6.1 Test Description.....................................................115
3.6.2 Specimen Preparation and Test Procedure..............................116
3.6.3 Measurement and Data Reduction.......................................118
3.6.4 Test Programs........................................................119
3.6.5 Test Results and Discussions.........................................122
3.7 Summary and Concluding Remarks........................................131
4. The SGP Test Apparatus.................................................133
4.1 First- and Second-Generations of SGP Test Apparatus....................134
4.1.1 First-Generation SGP Test Apparatus----------------------------------134
4.1.2 Second-Generation SGP Test Apparatus................................135
42 The Modified SGP Test Apparatus-----------------------------------------138
4.2.1 Apparatus Configurations---------------------------------------------138
4J2.2 Boundary Conditions__________________________________________________145
423 Specimen Preparation and Test Procedure--------------------------------149
4.2.4 Instrumentation_____________________________________________________154
viii


4.3 Test Programs.......................................................161
5. Behavior of GRS Mass subject to Monotonic Loading and
Unloading-Reloading Cycles and Finite Element Analysis..................166
5.1 Monotonic-Loading MSGP Test Results and Discussions................167
5.2 Unloading-Reloading MSGP Test Results and Discussions..............178
5.3 Effects of Preloading on Deformation and Strength of GRS Mass.......198
5.3.1 Effects of Preloading on Deformation..............................198
5.3.2 Effects of Preloading on Strength..................................203
5.4 Finite Element Analysis of the SGP Test.............................208
5.4.1 Program Description................................................208
5.4.2 Material and Interface Behavior Models.............................209
5.4.3 Determination of Model Parameters..................................223
5.4.4 Finite Element Modelling...........................................224
5.4.5 Comparison of Finite Element Analysis with SGP Test Results........230
5.4.6 Stresses in GRS Mass in the SGP test...............................235
5.5 Summary and Concluding Remarks.......................................242
6. The Simplified Preloading-Reloading Model for GRS Mass.............245
6.1 Load Transfer Module................................................245
6.1.1 Load-Transfer Analysis.............................................246
6.1.2 Comparison of Load-Transfer Analysis with
Experimental and Numerical Analysis Results of the APSR Test............258
6.1.3 Average Stresses in GRS Mass...................................... 267
63. Deformation Module...................................................268
6.2.1 Average Stress-Displacement Diagram________________________________268
6.2.2 Vertical and Horizontal Displacements______________________________272
6.3 Comparison of SPR Model Prediction with SGP Test Results............273
6.3.1 Application of SPR Model to SGP Test______________________________274
6.3.2 Determination of Material and Interface Properties----------------276
IX


633 Calculation Example................................................287
6.3.4 Comparison of SPR Model Prediction with SGP Test Results.........293
6.4 Parametric Study on Deformation of GRS Mass______________________303
6.5 Summary and Concluding Remarks.....................................308
7. Correlation between SGP Tests and Preloaded GRS Structures.........310
7.1 The FHWA Pier______________________________________________________311
7.1.1 Load Test Results................................................311
7.1.2 SGP Test Results.................................................312
7.1.3 Correlation between SGP Test and FHWA Pier.......................313
72 The Black Hawk Abutments............................................319
7.2.1 Load Test Results................................................319
722 SGP Test Results...................................................320
7.2.3 Correlation between SGP Test and Black Hawk Abutments____________320
7.3 Summary and Concluding Remarks...................................326
8. Summary, Conclusions, and Recommendations..........................327
8.1 Summary..........................................................327
8.2 Findings and Conclusions-------------------------------------------329
8.3 Recommendations....................................................333
References.............................................................335
x


FIGURES
Figure
2.1 Possible Stress-Paths in Triaxial Compreesion (After Lade and Duncan, 1976).. 9
2.2 Stress-Paths of Conventional Triaxial Compression
(After Lade and Duncan, 1976)................................................9
2.3 Stress-Strain Curve from Cyclic Triaxial Compression Tests
(After Holubec, 1968).......................................................... 10
2.4 Unloading-Reloading Modulus of Soil in Triaxial Compression
(After Duncan and Chang, 1970)...............................................10
2.5 Response of HPDE Geogrid Specimens to Multi-Increment and Single
Increment Cyclic Loading (After Barthurst and Cai, 1994).....................15
2.6 Charateristics of Cyclic Response of Geosynthetic Specimen
(After Barthurst and Cai, 1994)..............................................15
2.7 Area of Hysteresis Loops for HPDE and PET Specimens
(After Barthurst and Cai, 1994)..............................................16
2.8 Unloading-Reloading Modulus for HPDE and PET Specimens
(After Barthurst and Cai, 1994)..............................................16
2.9 Interface Friction Angle as Function of Number of Repeated Loading for
Ottawa Sand on HPDE (After ORouke et al., 1990).............................18
2.10 Increased Confinement Concept of Soil Reinforcement (After Yang, 1974) 21
2.11 Stress-Strain Relationships from Triaxial Compression Tests on
Reinforced Sand (After Gray and AI-Refeai, 1987).............................21
2.12 Ratcheting Mechanism (After Tatsuoaka et al., 1997)_____________________23
XI


2.13 Schematic Diagram of Preloaded/Prestressed GRS Structure
(After Tatsuoka et al., 1997)...............................................26
2.14 Typical Load Path of Preloaded/Prestressed GRS Structure
(After Tatsuoka et al., 1997)...............................................27
2.15 Principal Elements of FHWA Pier (After Adams, 1997)...................29
2.16 Preloading Assembly of FHWA Pier (After Adams, 1997)..................30
2.17 Prototype Preloaded/Prestressed GRS Bridge Pier
(After Uchimura et al., 1998)...............................................34
2.18 Black Hawk Abutments (After Wu et al. 1999)...........................35
2.19 Preloading Assembly of Black Hawk Abutments (After Wu et al., 1999)___36
2.20 Behavior of a Unit Cell With and Without Inclusions:
(a) Dense Sand; (b) Loose Sand (After McGown et al., 1978)..................42
2.21 Schematic Diagram of Plane Strain Compression Test Specimen
(After Tatsuoka and Yamauchi, 1986).........................................43
2.22 Plane Strain Compression Test Results for Unreinforced and
Reinforced Sand Specimens (After Tatsuoka and Yamauchi, 1986)...............44
2.23 Cross Section Through the APSR Cell (After Larson, 1992)..............45
2.24 Stress Distribution in a Steel Inclusion of the APSR Cell
(After Whittle et al., 1992)................................................46
2.25 Schematic Diagram of the Unit Cell Device:
(a) Profile and (b) Plan View Section (After Boyle, 1995)...................47
2.26 Unit Cell Device Test Results of Unreinforced and Reinforced Soils
(After Boyle, 1995)..........................................................48
3.1 Grain Size Distribution of Ottawa Sand..................................53
3 2. Grain Size Distribution of Road Base Soil..............................54
3.3 Moisture Content-Dry Unit Weight Relationship of Road Base Soil.........54
3.4 MTS-810 Loading System--------------------------------------------------59
3.5 General Loading Sequences_______________________________________________60
xn


3.6 Conventional Triaxial Compression Test Apparatus...................62
3.7 Test Results of Monotonic-Loading CTC Tests on Ottawa Sand
(Tests T-M-S1,2, and 3)................................................73
3.8 Test Results of Monotonic-Loading CTC Tests on Road Base Soil
(Tests T-M-RB1,2, and 3)...............................................74
3.9 Results of Test T-UR-S1 (Confining Pressure = 69 kPa)..............75
3.10 Results of Test T-UR-S2 (Confining Pressure = 207 kPa).............76
3.11 Results of Test T-UR-S3 (Confining Pressure = 345 kPa).............77
3.12 Results of Test T-UR-RB1 (Confining Pressure = 69 kPa).............78
3.13 Results of Test T-UR-RB2 (Confining Pressure = 207 kPa)............79
3.14 Results of Test T-UR-RB3 (Confining Pressure = 345 kPa)............80
3.16 Erl-z versus a^-PL Relationships of Ottawa Sand.................85
(o'. )/
3.17 Erl-ps versus Relationships of Ottawa Sand......................86
3.18 Erl-z verstis Relationships of Road Base Soil.............87
(o', -Oj)/
3.19 Erl-ps versus a^-PL Relationships of Road Base Soil.............88
(o',-Oj),
3.20 Deformation Modulus versus Confining Pressure Relationships
of Ottawa Sand..........................................................89
3.21 Deformation Modulus versus Confining Pressure Relationships
of Road Base Soil.......................................................90
322 p-q Diagram at Failure of Ottawa Sand..............................91
3.23 p-q Diagram at Failure of Road Base Soil__________________________92
322 Typar3301 Specimen.................................................97
323 LE Test Setup for Typar 3301 Specimen..............................98
3.24 Amoco 2044 Specimen................................................99
xiii


3.25 LE Test Setup for Amoco 2044 Specimen...............................
3.27 Tensile Load versus Axial Strain Relationship, Test G-M-3301........
3.28 Tensile Load versus Axial Strain Relationships, Tests G-M-2044-1,2...
3.29 Tensile Load versus Axial Strain Relationships of
Tests G-UR-3301-1,2, and 3 (Stress-Controlled Part).......................
3.30 Tensile Load versus Axial Strain Relationships of Test G-M-3301 and
Tests G-UR-3301-1,2,3 (Strain-Controlled Part)............................
3.31 Tensile Load versus Axial Strain Relationship
of Tests G-UR-2044-l, 2, and 3 (Stress-Controlled Part)...................
3.32 Tensile Load versus Axial Strain Relationships of Test G-M-2044-land
Tests G-UR-2044-l, 2 (Strain-Controlled Part)...............................
3.34 Stiffness Ratio versus
Relationships of Typar 3301 and
Amoco 2044
100
108
108
109
110
111
112
113
335 Failure Load Ratio versus
Relationships Typar 3301 and
Amoco 2044...................................................................114
3.35 Results of Tests DS-UR-(S+2044)-l, 2, and 3............................125
336 Results of Tests DS-UR-(RB+2044)-l, 2, and 3............................126
3.37 Interface Stiffness versus Normal Stress Relationships of Ottawa sand and
Amoco 2044 Interface.........................................................127
3.38 Interface Stiffness versus Normal Stress Relationships of Road Base Soil
and Amoco 2044 Interface.....................................................128
3.39 Peak Shear Stress versus Normal Stress Relationships of Ottawa sand
and Amoco 2044 Interface.....................................................129
3.40 Peak Shear Stress versus Normal Stress Relationships of Road Base Soil
and Amoco 2044 Interface.....................................................130
4.1 First-Generation SGP Test (After Wu and Helwany, 1996)__________________137
XIV


4.2 Second-Generation SGP Test (After Ketchart and Wu, 1996)................137
4.3 Schematic Diagram of the Modified SGP Test..............................139
4.4 The Modified SGP Test Apparatus on MTS-810 Loading System
before Testing.............................................................140
4.5 Specimen Dimensions of the Modified SGP Test............................141
4.6 Rigid Container of the Modified SGP Test Apparatus.......................146
4.7 Top View of the Modified SGP Test Apparatus.............................147
4.8 Cross Section of the Modified SGP Test Apparatus (Section A-A)..........148
4.9 Specimen Preparation for Ottawa Sand Specimen.........................156
4.10 Specimen Preparation for Road Base Soil Specimen.......................157
4.11 Strain Gage Layout...................................................158
4.12 Strain Gage Calibration with LE Test.................................159
4.13 Calibration Curve for Strain Gages...................................160
5.1 Vertical Load versus Displacement Relationships of Tests
P-M-RB and P-M-(RB+2044)...................................................169
5.2 Vertical Load versus Displacement Relationships of Tests
P-M-S, P-M-(S+3301), and P-M-(S+2044)......................................170
5.3 Horizontal Displacements of Points T, M, and B at 4 kN, 8 kN, and 11 kN.171
5.4 Failure Modes in SGP tests............................................174
5.5 Diagonal Shear Failure................................................175
5.6 Wedge-Type Shear Failure..............................................176
5.7 Vertical Load versus Displacement Relationships of Test P-UR-S........188
5.8 Vertical Load versus Displacement Relationships of
Test P-UR-(S+2044)-l.......................................................189
5.9 Vertical Load versus Displacement Relationships of
Test P-UR-(S+2044)-2.......................................................190
5.10 Vertical Load versus Displacement Relationships of
Test P-UR-(S+2044)-3.......................................................191
XV


I
5.11 Vertical Load versus Displacement Relationships of Test P-UR-RB..192
5.12 Vertical Load versus Displacement Relationships of
Test P-UR-(RB+2044)....................................................193
5.13 Vertical Applied Pressure versus Displacement Relationships of
the FHWA pier..........................................................194
5.14 Conceptual Stress Diagrams for the RL-Z and RL-PS Paths..........195
5.15 Vertical Load versus Reinforcement Strain Relationship of
Test P-UR-(RB+2044), Gage R-3..........................................196
5.16 Vertical Load versus Displacements of Virgin Specimen from
Test P-M-(S+2044) and Preloaded Specimens from Tests P-UR-(S+2044)-1,2,3... 200
5.17 Vertical Load versus Displacements of Virgin Specimen from
Test P-M-(RB+2044) and Preloaded Specimens from Tests P-UR-(RB+2044)......201
5.18 Vertical Load versus Reloading Displacement Relationships (RL-Z path)
of Tests P-UR-S and P-UR-(S+2044)-l.......................................204
5.19 Vertical Load versus Reloading Displacement Relationships (RL-Z path)
of Tests P-UR-RB and P-UR-(RB+2044).......................................205
5.20 Vertical Load versus Reloading Displacement Relationships (RL-PS path)
of Tests P-UR-RB and P-UR-(RB+2044).......................................206
5.21 Hyperbolic Model of Stress-Strain Behavior
(After Duncan and Chang, 1970)............................................221
5.22 Component of Interface Elements and Hyperbolic
Shear Stress-Relative Shear Displacement (After Clough and Duncan, 1969)..222
5.23 Calculated versus Measured CTC Test Results of Road Base Soil........226
5.24 Calculated versus Measured DS Test Results of Road Base Soil and
Amoco 2044 Interface......................................................227
5.25 Finite Element Discretizations of MSGP Test Specimens...............229
5.26 Measured and Calculated Vertical and Average Horizontal Displacements
ofTestP-M-RB..............................................................232
XVI


5.27 Measured and Calculated Vertical and Average Horizontal Displacements
of Test P-M-(RB+2044)....................................................233
5.28 Calculated Strain Distributions in the Middle Reinforcement Layer.234
5.29 Vertical Stress Distributions at 6 kN Vertical Load of
Tests P-M-RB and P-M-(RB+2044)...........................................238
5.30 Horizontal Stress Distributions at 6 kN Vertical Load of
Tests P-M-RB and P-M-(RB+2044)...........................................239
5.31 Shear Stress Distributions at 6 kN Vertical Load of
Tests P-M-RB and P-M-(RB+2044)...........................................240
5.32 Distribution of Minor Principal Stress Ratio at 6 kN Vertical Load of
Test P-M-(RB+2044).......................................................241
6.1 An Idealized Plane-Strain GRS Mass for the SPR Model................247
6.2 Equilibrium of Differential Soil and Reinforcement Elements
(Reproduced from Hermann and Al-Yassin, 1978)............................248
6.3 Schematic Diagrams of the APSR Cell (After Whittle et ai, 1991)....259
6.4 Predicted and Measured Normalized Reinforcement Stress Distributions
in the APSR Cell.........................................................264
6.5 Average Stress-Displacement Diagram for Monotonic Loading...........270
6.6 Average Stress-Displacement Diagram for Unloading and Reloading.....271
6.7 The SGP-GRS Mass....................................................275
6.8 Tensile Load versus Axial Strain Relationship of Amoco 2044
at Small Strain..........................................................284
6.9 Predicted versus Measured Vertical and Average Horizontal
Displacements of Test P-M-(S+2044).......................................294
6.10 Predicted versus Measured Vertical and Average Horizontal
Displacements of Test P-M-(RB+2044)......................................295
6.11 Predicted versus Measured Reloading Vertical and Average Horizontal
Displacements of Test P-UR-(S+2044), RL-Z path___________________________298
XVII


6.12 Predicted versus Measured Reloading Vertical and Average Horizontal
Displacements of Test P-UR-(S+2044), RL-PS path..........................299
6.13 Predicted versus Measured Reloading Vertical and Average Horizontal
Displacements of Test P-UR-(RB+2044), RL-Z path..........................300
6.14 Predicted versus Measured Reloading Vertical and Average Horizontal
Displacements of Test P-UR-(RB+2044), RL-PS 1 path.......................301
6.15 Predicted versus Measured Reloading Vertical and Average Horizontal
Displacements of Test P-UR-(RB+2044), RL-PS2 path........................302
6.16 Vertical Displacement Ratio versus Stiffness Ratio Relationships...306
6.17 Horizontal Displacement Ratio versus Stiffness Ratio Relationships.307
7.1 Vertical Applied Pressure versus Displacement relationships of
the FHWA Pier...........................................................315
7.2 Vertical Load versus Vertical Displacement Relationships of the SPG Test 316
7.3 Vertical Improvement Ratio versus Applied Load Level Relationships of
the FHWA Pier and SGP Test..............................................317
7.4 Horizontal Improvement Ratio versus Applied Load Level Relationships of
the FHWA Pier and SGP Test..............................................318
7.5 Vertical Applied Pressure versus Average Settlement Relationships
of Footing #1 of the Black Hawk Abutments................................322
7.6 Vvertical Applied Pressure versus Average Settlement Relationships
of Footing #3 of the Black Hawk Abutments................................323
7.7 Predicted and Measured Vertical Load versus Vertical Displacement of
the SGP Tests...........................................................324
7.8 Vertical Improvement Ratio versus Applied Load Level Relationships of
the Black Hawk Abutments and SGP Tests...................................325
xvm


!
TABLES
Table
2.1 Summary of Four Plane Strain Compression Tests for Reinforced-Soil Mass... 49
3.1 Some Index Properties of Goesynthetics................................55
3.2 CTC Test Program for the Ottawa Sand.................................68
3.3 CTC Test Program for the Road Base Soil..............................69
3.4 LE Test Program for Typar 3301......................................102
3.5 LE Test Program for Amoco 2044......................................103
3.6 DS Test Program for the Ottawa Sand and Amoco 2044 Interface........120
3.7 DS Test Program for the Road Base Soil and Amoco 2044 Interface.....121
4.1 SGP Test Program for the Ottawa Sand Specimen........................163
4.I(Cont.) SGP Test Program for the Ottawa Sand Specimen.................164
4.2 SGP Test Program for the Road Base Soil Specimen....................165
5.1 Summary of Failure Loads of the SGP Tests............................207
5.2 Summary of Hyperbolic Soil Parameters for Finite Element Analysis...225
5.3 Summary of Interface Properties for Finite Element Analysis..........225
6.1 Input Parameters for Load Transfer Analysis of the APSR Test Specimen
(Data from Abramento, 1993)..............................................263
6.2 Reference Properties for Comparison of Maximum Normalized
Reinforcement Stress for Table 6.3 (Data from Whittle et al., 1991)......265
6.3 Comparison of Maximum Normalized Reinforcement Stress from
Finite Element Analysis and Load-Transfer Analysis.......................266
6.4 Summary of Soil Properties for the SPR Model.........................281
6.4(Cont.) Summary of Soil Properties for the SPR Model..................282
XIX


6.5 Summary of Interface Properties for the SPR Model..........................286
6.6 Calculation Example of the SPR Model for SGP-GRS Mass......................289
6.6(Cont.) Calculation Example of the SPR Model for SGP-GRS Mass...........290
6.6(Cont.) Calculation Example of the SPR Model for SGP-GRS Mass...........291
6.6(Cont) Calculation Example of the SPR Model for SGP-GRS Mass...........292
6.7 Properties for the Baseline GRS Mass......................................305
XX


1. Introduction
1.1 Problem Statement
A geosynthetic-reinforced soil (GRS) mass is a soil mass containing
horizontally placed layers of geosynthetic reinforcement When subject to a vertical
load, a GRS mass generally exhibits higher stiffness and higher load carrying
capacity than a soil mass without reinforcement The increase in stiffness and
strength results from an internal restraining effect imposed by the geosynthetic
reinforcement on the GRS mass. The geosynthetic reinforcement restrains
deformation of the GRS mass along the axial direction of the reinforcement due to
soil-geosynthetic interface friction.
It is known that some deformation of a GRS mass is needed before sufficient
tensile force in the reinforcement is developed to mobilize the internal restraining
effect in the GRS mass. The required deformation of a GRS mass may be
unacceptable in some applications such as bridge piers and abutments. An effective
means to limit the deformation of a GRS mass under load applications is to preload
the GRS mass, as suggested by Tatsuoka et al. in 1997.
The preloading technique involves applying a fairly large load on the GRS
mass, followed by unloading to either a lower load level or zero. The GRS mass will
I


typically be stiffer in a subsequent reloading path because: (a) the soil stiffness is
increased after it has been preloaded, and (b) the internal restraining effect has been
mobilized during the preloading stage.
A number of full-scale tests have been conducted to examine the effects of
preloading on different GRS structures, including geogrid-reinforced soil retaining
walls (Tatsuoka et al., 1997), a geotextile-reinforced soil bridge pier (Adam, 1997), a
geogrid-reinforced soil bridge pier (Uchimura et al., 1998), and geotextile-reinforced
soil bridge abutments (Ketchart and Wu, 1998). These full-scale tests demonstrated
the feasibility of preloading GRS structures in actual construction. They also
provided useful data on the performance of preloaded GRS structures. However, the
fundamental behavior of preloaded GRS masses has not yet been fully elucidated.
Specifically, two important questions remain unanswered.
1. What is the appropriate loading magnitude and loading sequence to effectively
preload a GRS mass?
2. How much deformation is to be expected in a preloaded GRS mass?
It is cost prohibitive and time consuming to seek answers to these questions
by conducting a long list of the full-scale tests with different types of soils and
geosynthetics under various loading magnitudes and loading sequences.
This study was undertaken to provide answers to the two questions. An
extensive experimental study was conducted on a number of soils, geosynthetics, soil-
geosynthetic interfaces, and GRS masses under various loading sequences. A
i
I
j
i
2


modified soil-geosynthetic performance (SGP) test apparatus was devised for
investigating the behavior of GRS masses. The results of the experimental study
provide answers to the first question. To answer the second question, a simplified
analytical model, referred to as the simplified preloading-reloading (SPR) model, was
developed for estimating deformation of a preloaded GRS mass.
1.2 Research Objectives
The objectives of this study were two-fold. The first objective was to gain a
better understanding of the behavior of preloaded GRS masses comprising different
types of soils and geosynthetics and subjected to various loading sequences. The
second objective was to develop a simplified analytical model for estimating
deformation of a preloaded GRS mass.
13 Method of Research
To achieve the research objectives, the following tasks were taken in this
study:
Task 1: Review previous studies on the behavior of sands, geosynthetics, soil-
geosynthetic interfaces, and GRS masses subject to unloading-reloading cycles, and
on plane-strain tests for reinforced soil masses (Chapter 2).
Task 2: Conduct laboratory tests to examine the behavior of different soils,
geosynthetics, and soil-geosynthetic interfaces subject to unloading-reloading cycles.
The test programs included conducting conventional triaxial compression tests on
3


soils, in-isolation load-extension tests on geosynthteics, and direct shear tests on soil-
geosynthetic interfaces. (Chapter 3)
Task 3: Devise a test apparatus and establish test procedures for investigating the
behavior of GRS masses subject to different loading sequences. (Chapter 4)
Task 4: Conduct laboratory tests to investigate the behavior of GRS masses subject to
different loading sequences. Finite element analyses was conducted to examine the
stress distribution in a GRS mass in the SGP test. (Chapter 5)
Task 5: Develop a simplified analytical model for estimating deformation of a
preloaded GRS mass. (Chapter 6)
Task 6: Evaluate correlations between SGP test and preloaded GRS structures
(Chapter 7).
4


2. Literature Review
A review of some previous studies on the behavior of soils, geosynthetics,
soil-geosynthetic interfaces, and GRS masses subject to unloading-reloading cycles is
presented in this chapter. Such the unloading-reloading cycles are categorized as a
static load based on the definition by Ishihara (1998), as the load application lasts for
more than ten seconds. In addition, the preloaded GRS structures are briefly
described. This chapter also presents a review of four plane strain tests conducted on
reinforced soils.
2.1 Behavior of Sand subject to
Unloading-Reloading Cycles
When a mass of sand is subjected to a stress variation, its deformation can be
considered as the sum of a recoverable (elastic) component and an irrecoverable
(plastic) component. From the standpoint of the deformation of grains and sliding
between grains, the recoverable part is due to the elastic deformation of individual
grains, whereas the irrecoverable is primarily caused by the sliding between
individual grains.
Lade and Duncan (1976) proposed criteria to define primary loading,
unloading, and reloading modes for different stress-paths of a triaxial compression
test. Figure 2.1 shows a diagram representing the stress paths which can be produced
5


in a triaxial compression in terms of the deviator stress (03-03) and the confining
stress (03). A stress level is used as the basis in formulating a criterion for the
mode of deformation. The stress level refers to the fraction of the soil strength that
is mobilized. For a cohesionless soil, a straight line passing through the origin of ( 03) versus <73 diagram represents a constant stress level. Proportional loading occurs
when the stresses change in a manner that the stress level remains constant (stress-
paths 5 and 9). Unloading is experienced whenever the stress level decreases (stress-
paths 6,7, 8, and 11). Reloading is said to occur whenever the stress level increases
but remains less than the past maximum value experienced by the soil (stress-path
10). Primary loading is experienced only when the stresses change in such a manner
that the stress level exceeds its past maximum value (stress-paths 1,2,3, and 4). The
stress-path for a conventional triaxial compression test in which the confining
pressure remains constant while the axial stress is increased is represented by a
vertical line as shown in Figure 2.2.
When a soil specimen is unloaded, individual grains do not rebound to their
original positions but remain approximately in their displaced positions (Makhlouf
and Stewart, 1965). If a soil specimen is unloaded from a stress state, A, (see Figure
2.3) to another stress state, B, then reloaded again to the original stress condition, A,
along the same stress-strain curve, the unloading and reloading stress-strain paths
coincide in a reversible process (Holubec, 1968). In general, the identity of the
6


unloading and reloading paths is not perfect, especially in the high-stress range, as
evidenced by a hysteresis loop. A hysteresis loop exists as shown in the third
unloading-reloading cycle of the stress-strain curve in Figure 2.3. The hysteresis loop
in the unloading-reloading cycle implies that: (a) there is no longer a one-to-one
relationship between stress and strain in this unloading-reloading region, and (b)
energy is dissipated in an unloading-reloading cycle, which also implies inelastic
response (Wood, 1990).
Holubec (1968) suggested that the identity of the unloading and reloading
paths can be assumed if the width of a hysteresis loop is small compared with the
magnitude of the reversible strains, or when specimens are unloaded to zero-shearing
stress from a stress less than approximately 80% of the maximum deviator stress.
Barden et al. (1969) also observed that if the unloading-reloading cycle takes place
when a principle stress ratio (01/03) less than two-third of the peak value, a hysteresis
loop is small. However, if the unloading-reloading cycle is in an region which has
peak or post-peak values of principle stress ratios, then the hysteresis loop is
significant. Note that the width of the hysteresis loop of sand in a conventional
triaxial compression test was the greatest in the first cycle and decreased in
subsequent cycles (Makhlouf and Stewart, 1965).
The deformation of sand in the unloading-reloading range which takes place
at moderate stress levels (r.e. not close to the failure stress) can be approximately
characterized as linear elastic (Holubec, 1968; Duncan and Chang, 1970; Coon and
7


Evans, 1971; Lade and Duncan, 1975). The average secant modulus of the
unloading-reloading loop was defined as the unloading-reloading modulus (Eur) by
Duncan and Chang (1970) as shown in Figure 2.4. The unloading-reloading modulus
is proportional to the confining stress (Duncan and Chang, 1970). The unloading-
reloading modulus depends upon the change of the de viator stress in an unloading-
reloading cycle (Makhlouf and Stewart, 1965). The unloading-reloading modulus
increases if the change of deviator stress is held constant but the magnitude of
minimum deviator stress is increased. With a constant magnitude of the maximum
stress, the unloading-reloading modulus decreases with decreasing minimum deviator
stress in unloading-reloading cycles.
The deformations of granular soils in the primary loading are almost
unaffected by the previous unloading-reloading cycles which occur at lower stress
levels (Makhlouf and Stewart, 1965; Ko and Scott, 1967). Unlike those in the
primary loading stress-path, the deformation of sand under a reloading stress path is
very much dependent on the stress histories it has experienced.
S


Figure 2.1: Possible Stress-Paths in Triaxial Compreesion (After Lade and
Duncan, 1976)
Figure 22: Stress-Paths of Conventional Triaxial Compression (After Lade
and Duncan, 1976)
9


Figure 23: Stress-Strain Curve from Cyclic Triaxial Compression Tests
(After Hoiubec, 1968)
Figure 2.4: Unloading-Reloading Modulus of Soil in Triaxial Compression
(After Duncan and Chang, 1970)
10


Yoshimi et al. (1975) used an adaptation of a quicksand tank to reproduce
uniform void ratios in large samples of sand using a controlled upward flow of water.
By reversing the direction of water flow, one-dimensional loading was induced over
the entire sample. They found that a normally consolidated sand sample was about
six times more compressible than a prestressed sample, even though their initial void
ratios or densities were equal.
Lambrechts and Leonards (1978) conducted a series of triaxial compression
tests under different stress paths to examine effects of stress history on deformation of
sand. Each set of stress paths used in simulating different stress histories was a
combination of stress path segments including proportional loading, unloading, and
reloading. At the end of each set of such stress paths, the axial stress was increased
while maintaining a constant confining pressure as in a conventional triaxial test
They found that by prestressing the sand under Ko-condition, the modulus of
deformation under the conventional triaxial compression loading increased by one
order of magnitude.
Bishop and Eldin (1953) studied the effect of stress history on the angle of
internal friction of sand by conducting a number of triaxial compression tests. They
concluded that the angle of internal friction of sand is independent of the stress
history. This conclusion was confirmed by Lade and Duncan (1976) and Lambrechts
and Leonards (1978).
II


Based on the literature review outlined above, the behavior of sand subject to
preloading-reloading loads is summarized as follows:
1) Elastic behavior can be assumed for sand under unloading-reloading cycles
that take place at moderate stress levels.
2) A hysteresis loop exists in unloading-reloading cycles. The hysteresis loop
indicates inelastic behavior and energy dissipation during unloading and reloading.
The width or area of the hysteresis loop becomes significant during unloading and
reloading at high stress levels.
3) An average secant modulus of the unloading-reloading loop can be
represented by the unloading-reloading modulus (Eur). The unloading-reloading
modulus (Eur) is proportional to the confining stress and also depends on maximum
and minimum values of deviator stress change in the unloading-reloading cycles.
4) Deformation of sand under a reloading stress path is strongly influenced by
the stress history. The deformation moduli increase significantly after the sand has
been prestressed.
5) The angle of internal friction or the shear strength of sand is independent of
the stress history.
12


2.2 Behavior of Geosynthetics subject to
Unloading-Reloading Cycles
Some studies have been conducted by in-isolation cyclic load extension tests
to examine the cyclic behavior of geosynthetics. Barthurst and Cai (1994) conducted
a series of in-isolation cyclic load-extension tests on HDPE (High Density
Polyethylene) and PET (Polyester) geogrid specimens. The specimens were tested at
different loading frequencies from 0.1 to 3.5 Hz and over a range of load amplitudes.
Figure 2.5 shows typical load-strain response curves of the HPDE geogrid specimens
under multi-increment and single increment cyclic loadings. A hysteresis loop exists
at all unloading-reloading cycles. Accumulative plastic strains due to multiple cycles
of cyclic loading are evident. Some qualitative features of a cyclic load-deformation
response curve are illustrated in Figure 2.6. Figure 2.6 identifies the parameters that
can be used to characterize the load-deformation response as a function of strain. A
non-linear hysteresis load-deformation loop for each unloading-reloading cycle (sur,
Tut) is defined by the average unloading-reloading modulus (Jur) of the unloading-
reloading cycle and its contained area (Ayr).
The area of a hysteresis loop (Ayr) of the cyclic load-deformation curves of the
geogrid specimens was found to be strongly influenced by the strain level and the
frequency of loading. The area, Aur, increases with the strain level and decreases
with increasing frequency at a given strain, as shown in Figure 2.7. It should be
noted that below 0.5% strain of the HDPE geogrid and 0.8% strain of the PET geogid,
13


the specimens behaved in a linear elastic manner with fully recoverable strain. Figure
2.8 shows the average unloading-reloading modulus versus strain relationships for
different load amplitudes and frequencies. The average unloading-reloading
modulus, Jut, of the HDPE specimens reduces with the strain level, whereas the PET
specimens showed a reduction of Jur up to about 3% of strain and followed by an
increase.
Similar in-isolation cyclic load-extension tests on HDPE geogrids were
conducted by Nocola and Montanelli (1997). The specimens were tested at different
loading frequencies from 0.1 to 1.0 Hz and over different cyclic loading ranges. They
have found that the unloading-reloading modulus, Jur, increases with strains until it
reaches a yield point, after which the unload-reload tensile modulus gradually
decreases with increasing strains.
In summary, the unloading-reloading behavior of geosynthetics can be
quantified by the unloading-reloading modulus (Jur) and the area of a hysteresis loop
(Aur). The hysteresis loop occurs when geosynthetics are subjected to unloading-
reloading cycles. The area of hysteresis loop increases with strain level. At small
strains (0.5% to 0.8%), the area of hysteresis loop becomes negligible, and the
geogrids behave in a linear manner. For HPDE geogrids, the unloading-reloading
modulus increases slightly with increasing strains until it reaches a yield point, after
which the unloading-reloading modulus reduces with increasing strains.
14


Figure 2.5: Response of HPDE Geogrid Specimens to Multi-Increment and
Single Increment Cyclic Loading (After Barthurst and Cai, 1994)
average unload-reload stiffness Jur
/
t
e(%)
Figure 2.6: Charateristics of Cyclic Response of Geosynthetic Specimen
(After Barthurst and Cai, 1994)
15


Figure 2.7: Area of Hysteresis Loops for HPDE and PET Specimens (Alter
Barthurst and Cai, 1994)
1 imdi tocwnwnt cmdh toad tost*
Figure 2.8: Unloading-Reloading Modulus for HPDE and PET Specimens
(After Barthurst and Cai, 1994)
16


23 Behavior of Soil-Geosynthetic Interfaces subject
to Unloading-Reloading Cycles
A limited number of works on the behavior the soil-geosynthetic interfaces
subject to unloading-reloading cycles were available in the literature. ORourke et al.
(1990) conducted a series of direct shear tests on Ottawa sand and high-density
polyethylene (HDPE) geosynthetic. They have found that the shear strength of the
interface was not affected by the repeated loading as shown in Figure 2.9. Figure
2.10 show the shear strength of the interface plotted versus number of repeated
loading before shear to failure.
2.4 Behavior of GRS Masses subject to
Unloading-Reloading Cycles
2.4.1 General Behavior
A geosynthetic-reinforced soil (GRS) mass is a soil mass embedded with
layers of geosynthetic reinforcement. In this study, unless otherwise specified, the
reinforcement layers are horizontally oriented. This section begins with a
presentation of the strength and deformation behavior of a GRS mass and followed by
the effects of preloading on a GRS mass.
17


friction Anglo, 8
Figure 2.9: Interface Friction Angle as Function of Number of Repeated
Loading for Ottawa Sand on HPDE (After ORouke et al., 1990)
18


Under vertical loading, a GRS mass shows a higher load carrying capacity
than a soil mass without reinforcement This reinforcing effect of reinforcement has
been explained by an increased confinement concept by Yang (1974). The concept is
illustrated by Mohr stress diagram shown in Figure 2.10. The vertical and lateral
stresses are assumed to be major and minor principal stresses, respectively. Circle A
represents an at-failure stress state of a soil mass without reinforcement The vertical
and lateral stresses at failure for the soil mass is at and a3C, respectively (see Figure
2.10). With a reinforcement, assuming no slippage at the soil-reinforcement interface
and failure of the reinforced soil mass is due to rupture of the reinforcement, the
lateral stress at failure is increased by ACT3R, which is equal to the tensile strength of
the reinforcement. As a consequence, the vertical stress at failure increases to ctir,
i.e. a higher load carrying capacity is obtained.
Under a vertical load, the GRS mass exhibits both lateral and vertical
deformation responses. The soil expands laterally with the geosynthetic and
mobilizes tensile forces in the geosynthetic through the friction between the soil and
the geosynthetic. The tensile force in the geosynthetic restrains the lateral movement
of the soil and, consequently, reduces the vertical deformation.
The effect of reinforcement in reducing deformation of a soil mass can be
illustrated by triaxial compression test results conducted on unreinforced and
reinforced soil samples by Gray and AI-Refeai (1987) as shown in Figure 2.11.
19


Figure 2.11 shows that the stiffness or tangent moduli of the unreinforced and
reinforced specimens are almost the same until 1.5% of axial strain. In another word,
the internal restraining effect by the geosynthetic reinforcement is insignificant at
small strains. This is because the geosynthetic reinforcement requires some
deformation in order to mobilize the sufficient tensile force in the reinforcement.
Figure 2.11 also shows that at small strains (0 to 1.5%), the stiffness of a
reinforced soil is somewhat smaller than that in the unreinforced soil. Similar
behavior have been reported in triaxial compression tests by Brom (1977). Wu
(1989) has investigated this effect and concluded that the loss of compressive
stiffness in the reinforced soil is due to compression of the reinforcement itself. The
effect of compressibility of the reinforcement is pronounce in the triaxial tests
because ratios of the reinforcement spacing to the reinforcement thickness in the
triaxial tests are relatively small. The loss of stiffness at the small strains due to the
compressibility of the reinforcement is negligible in field construction because ratios
of the reinforcement spacing to the reinforcement thickness are much greater than
those in the triaxial tests.
20


Figure 2.10: Increased Confinement Concept of Soil Reinforcement (After
Yang,1974)
Figure 2.11: Stress-Strain Relationships from Triaxial Compression Tests on
Reinforced Sand (After Gray and AI-Refeai, 1987)
21


Deformation of a GRS mass is of major concern when it is to be used in
critical structures such as bridge piers and abutments. In order to limit the
deformation of a GRS mass, a preloading concept is applied to increase the stiffness
of the GRS structure (Tatsuoka et al., 1997). The preloading technique on a GRS
mass takes advantage of the fact that soil stiffness is increased after it has been
preloaded or prestressed. The preloaded GRS mass is also expected to behave nearly
elastically in a reloading path similar to what has been observed in a preloaded soil.
Preloading also mobilizes tensile strains in the geosynthetic reinforcement in a
service conditionthe so-called ratcheting mechanism (Tatsuoka et al., 1997). A
simple model of a soil-geosynthetic composite shown in Figure 2.12 illustrates this
mechanism. Under an applied pressure, ov, lateral deformation of the composite
occurs and results in a tensile force in the reinforcement. Upon unloading, most of
the lateral deformation of the soil does not rebound back. As a result, the
reinforcement has been stretched and the tensile strains are mobilized. This
mechanism also helps eliminating wrinkles that often occur during field placement of
geosynthetic layers.
22


Remforenwn;
a)
b>
Soil ^
Soil
11 rv
r*** \

i i H-r
TTT
e) >
c
*+* Tr
Figure 2.12: Ratcheting Mechanism (After Tatsuoakct al., 1997)
23


2.4.2 Preloaded GRS Structures
Since 1977, the preloading concept has been applied on four GRS structures.
They are:
1. Preloaded/Prestressed GRS walls in University of Tokyo, Tokyo, Japan
(Tatsuoka et al, 1997);
2. Preloaded GRS pier in Tumer-Fairbank Highway Research Center,
McLean, Virginia, USA (Adams, 1997), referred to as the FHWA pier;
3. Preloaded/Prestressed GRS bridge pier in Fukuoka city, Japan
(Uchimura et al., 1998);
4. Preloaded GRS bridge abutments in Black Hawk, Colorado, USA
(Wu et aL, 1999), referred to as the Black Hawk abutments.
2.4.1.1 Preloaded/Prestressed GRS Walls
Tatsuoka et al. (1997) proposed a new construction protocol, so-called
preloaded/prestressed (PL/PS) reinforced soil. The main purpose was to make
deformation of a GRS mass being nearly elastic and having very high stiffness under
applied loads.
A schematic diagram of a PL/PS GRS structure is shown in Figure 2.13.
Large preloading is applied by introducing tension into metallic tie rods that are
intruded through the reinforced soil mass and fixed to the bottom reaction block. The
tensile force in the tie rods and the corresponding compressive load in the backfill soil
24


function as prestressing to maintain the vertical confining pressure and results in high
stiffness in the vertical direction.
A typical PL/PS loading path involves preloading, sustained loading,
unloading to a desired prestress loading level, and reloading as shown in Figure 2.14.
A vertical load is applied up to a stress level, b, and sustained for a period of time.
After allowing the creep deformation during the preloading stage to occur (b to b')>
the load is reduced from b' to c as unloading. The stress level, c, is defined as an
initial prestress level. The vertical deformation is maintained constant at the stress
level, c, and, consequently, the prestress level decreases from c to c' due to plastic
deformation of the GRS mass known as the stress relaxation. A reloading stress path
is taking place from the stress level c' to d.
Full-scale loading tests of 5.4-m high geogrid-reinforced walls were
conducted at the University ofTokyo, Japan, to validate the PL/PS concept (Tatsuoka
et al., 1997). The full-scale test results showed that the stiffness of the soil mass with
compressive prestress was higher than that of the soil mass without prestress.
Deformation of soil after preloading was nearly elastic for a relatively small load
increment (Tatsuoka et al., 1997).
25


Figure 2.13: Schematic Diagram of Prcloaded/Prestressed GRS Structure
(After Tatsuoka et aL, 1997)
26


Creep deformation
/ during preloading
b'
Unloading
L- Reduced rate of
stress relaxation
Nearly elastic behaviour
during reloading
at a relatively fast rate
------------------ £
Figure 2.14: Typical Load Path of Preloaded/Prestressed GRS Structure (After
Tatsuoka et al., 1997)
27


2.4.1.2 FHWAPier
Detailed description of the FHWA pier has been presented by Adams (1997).
A brief description of the project is given below. The GRS pier was 5.4 m high with
base and top dimensions of 3.6m x 4.8m and 3.06m x 4.26m, respectively. The pier
was constructed with a well-graded gravel (GW-GM per ASTM D2487) and
reinforced with layers of geotextile sheets. The maximum dry density of the backfill
was 24 kN/m3 and the optimum moisture content was 5.0%, per AASHTO T180.
The average backfill density from nuclear density tests was 22.8 kN/m3. The
reinforcement was a high strength woven prolypropylene geotextile, Amoco 2044.
The vertical spacing of reinforcement was 0.2 m. Split face concrete (cinder) blocks,
of dimensions 0.2m x 0.2m x 0.4m, were dry-stacked to form the facing. The front
edge of each reinforcement sheet was placed between vertically aligned blocks to
achieve a frictional connection between the reinforcement layer and the facing blocks.
A schematic diagram of the pier is shown in Figure 2.15.
The loading mechanism of the GRS pier comprised hydraulic jacks and a
specially designed reaction system, as shown in Figure 2.16. The reinforced soil
mass was sandwiched between the top and bottom concrete pads which were
connected together with vertical steel rods. The hydraulic jacks were placed between
the top concrete pads and the reaction frame. Upon applying pressure to the
hydraulic jacks, the GRS pier was squeezed between the top and bottom pads.
28


Figure 2.15: Principal Elements of FHWA Pier (After Adams, 1997)
29


Lam&tA irmljly
Figure 2.16: Preloading Assembly of FHWA Pier (After Adams, 1997)
30


2.4.13 Preloaded/Prestressed GRS Walls
A prototype 2.7-m Mgh PL/PS geogrid-reinforced soil bridge pier, as shown
in Figure 2.17, was constructed at Fukuoka city, Japan, to support temporary railway
girders. It has been opened to service since the summer of 1997. Behavior of the
prototype PL/PS bridge pier during and after construction and in service was reported
by Uchimura et al (1998). The prototype pier showed very small transient and long-
term deformations compared to a nearby geogird-reinforced bridge abutment
constructed without preloading/prestressing subject to the same transient load from a
locomotive (Uchimura et al,1998).
2.4.1.4 Black Hawk Abutments
Detailed description of the Black Hawk abutments has been presented by Wu
et al (1999). A brief description of the project is given below. The abutments were
constructed in the mountain terrain above the city of Black Hawk, Colorado to
support a 36-m span steel arched bridge. The abutments were constructed with the
on-site soil (the Road Base soil used in the SGP test in this study) and reinforced with
layers of a woven geotextile (Amoco 2044) having a vertical spacing of 0.3 m.
Material properties of the soil and the reinforcement were given in Chapter 3. The
facing was of rock-faced type. The wall face was built by tightly stacking the rocks in
rows about 03 m in height. The front edge of each reinforcement sheet was placed
31


between vertically aligned rocks at the wall face to form a frictional connection
between the reinforcement layers and the facing rocks.
A series of sketches illustrating the geometry of the GRS abutments are in
Figure 2.18. As shown in Figure 2.18, each GRS abutment comprised a two-tier
rock-faced GRS mass, two square footings (on the GRS base mass), and a strip
footing (on the upper-tier GRS mass). Each abutment was constructed into a
mountain slope on opposite sides of a stream valley with a silty stream deposit. The
thickness of the silty soil layer was variable and considerably greater on the down
slope side of the mountain. The slopes were excavated to remove the silty soil, which
was considered unsuitable to support the abutments. The GRS abutments were
supported on a stiff soil layer underneath the silty soil layer.
As viewed from the faces (due east and west) (see Figure 2.18), the base of
the GRS mass was located at different depths of the excavated stiff soil. The variable
thickness of the GRS base mass was between 1.5 m to 7.5 m for the east abutment
and 1.5 m to 4.5 m for the west abutment. The width at the base of the GRS base
mass was 5.5 m. The lower part of the GRS base mass was embedded in the ground,
and the upper part was above the ground. Only the portion above ground was
constructed with rock facing. The height of the rock-faced wall varied from 1.0 m to
5.4 m in the east abutment and 1.0 m to 2.7 m in the west abutment. The upper-tier
GRS mass was perched on the backside of each GRS base mass. The upper-tier GRS
mass was 1.8-m thick and constructed in the same fashion as the GRS base mass.
32


The four square footings had a base area of 2.4 m x 2.4 m. The footing
thickness was about 1.65 m. The final thickness depended on the amount of
settlement due to preloading. The square footings on the west abutment were referred
to as Footing #1 (FI) and #4 (F4). The square footings on the east abutment were
referred to as Footing #2 (F2) and #3 (F3). The design load for each square footing
was 865 kN, equivalent to a vertical pressure of 150 kPa.
As shown in Figure 2.19, the preloading assembly for each footing consisted
of four 534 kN hollow-cored jacks ganged together with a manifold and connected to
a hydraulic electric pump. Each jack was placed on top of the square footing and
connected to a threaded rod by inserting the rod through the core of the jack. The
jack was sandwiched between the square footing and the steel bearing plates capped
with a nut threaded on the rod. On two jacks, 890 kN load cells were inserted
between the steel bearing plate and the nut. Installation of the threaded rods occurred
after construction of the GRS base mass. A survey located the perimeter of the
square footings and four prescribed points within the perimeter of the footing. At the
points, a reticulating air-percussion rotary drill rig bored 90-mm diameter holes
through the GRS mass, the stiff soil layer and into the underlying bedrock. The bond
length was about 3.5 m within the bedrock.
33


ft)
** Suttai
b)
Figure 2.17: Prototype Preloaded/Prestressed GRS Bridge Pier (After
Uchimuraetal., 1998)
34


I
!
-550-
East Abutment
550-
Dimension* or* .n
West Aoutment metre
East Abutment (due test)
West Abutment (due West)
Figure 2.18: Black Hawk Abutments (After Wit et al., 1999)
35


i I" "xoouk ,oc
TT l Souar* foct.rwj
i ,rJ i thread Sur
I !
Figure 2.19: Preloading Assembly of Black Hawk Abutments
36


To preload the GRS mass and the stiff soil layer underneath the footings,
hydraulic oil was pumped into the hydraulic jacks. As the cylinders advanced, the
GRS and the stiff soil were preloaded or squeezed between the footing and the
bedrock. After the preloading, each borehole was sealed with a grout mix.
2.5 Plane Strain Tests of Reinforced-Soil Mass
The behavior of reinforced-soil has been studied by using triaxial and plane
strain compression tests. Strictly speaking, triaxial compression test is only
applicable to a soil mass beneath the center line of a circular footing subject to
vertical and concentric loads. Most GRS structures (e.g., retaining walls and
embankments) are closed to being in a plane strain condition. Moreover, in typical
GRS structures, the geosynthetic reinforcement layers are placed with its stronger
direction perpendicular to the longitudinal direction of the plane strain structure. For
example, in GRS retaining walls, the stronger direction of a woven geotexdle
reinforcement is usually arranged to be perpendicular to the wall facing. Therefore,
plane strain compression tests generally give a better simulation of actual GRS
structures than triaxial compression tests.
Four plane strain tests conducted on the reinforced-soil masses by McGown et
aL (1978), Tatsuoka and Yamauchi (1986), Whittle et aL (1992) and Boyle (1995) are
reviewed in the following paragraphs.
McGown et aL (1978) employed a plane strain compression test apparatus to
study the effect of inclusion properties on the behavior of sand. The specimens were
37


Leighton Buzzard sand with and without inclusions of aluminium foil, aluminium
mesh and a non-woven melt bonded hetrofilament fabric. Specimen dimensions were
102 mm long, 102 mm high and 152 mm deep (i.e., in the longitudinal direction).
The apparatus had rigid lubricated top and bottom platens. The plane strain
condition was imposed by using two rigid lubricated side platens which were bolted
across the 102 mm xl02 mm faces. The confining pressure was applied using
vacuum and was kept constant during the tests. The test results were analyzed in
terms of the vertical stress-strain relationships and the internal deformations measured
by the stereo-viewing photogrammetric technique (Butterfield et at., 1970).
The behavior of the sand reinforced with the extensible and inextensible
inclusions is shown in Figure 2.20. The figures show the relationships of the
principal stress ratio (Gi/03) versus axial strain of loose and dense sands with and
without the inclusions. It is shown that the sands with the extensible inclusions were
more ductile than those with the inextensible inclusions. They concluded that the
overall load-deformation behavior of the reinforced soil system was significantly
influenced by the stiffness or the relative extensibility of the tensile inclusions.
Tatsuoka and Yamauchi (1986) conducted plane strain compression tests on
reinforced Toyoura sand specimens. The specimen dimensions were 80 mm wide, 75
mm high and 20 mm deep (r.e., the longitudinal direction), as shown in Figure 2.21.
The top and bottom sides of the specimen were lubricated. The side walls restraining
deformation of the specimen were also lubricated.
38


The reinforcement materials were brass plates, non-woven geotextiles, and
different types of rubbers. The average principal stress difference (o'1/0*3) versus
average minor principal strain (S3) relationships of the soil are shown in Figure 2.22.
The stress-strain relationships are similar to those of the plane strain tests conducted
by McGown et al. (1978) (of Figure 2.20), in that the sand reinforced with stiff
materials (brass plates) was more brittle than the sand reinforced with relatively less
stiff materials (geotextiles and rubbers). The test results also indicated that, in order
to mobilize a sufficient degree of tensile restraint in the composite, the non-woven
geotextiles required a larger soil deformation in the reinforcement direction than the
stiffer reinforcement materials.
Whittle et al. (1992) devised an Automated Plane Strain Reinforcement
(APSR) cell to study load transfer characteristics at working load levels of a
reinforced-soil mass. Figure 2.23 shows a schematic diagram of the APSR cell. The
soil specimen has dimensions of570 mm high, 450 mm wide, and 150 mm deep (i.e.,
the longitudinal direction). The major principal stress (cri< 500 kPa) was applied
through two pressurized water bags mounted on moveable rigid platforms. A
uniform lateral confinement (o3< 50 kPa) was provided by air-pressure. The
maximum tensile stress in the reinforcement was measured at the end which was
connected to a load cell. The stress in the reinforcement was induced by the stress
developed in the confining soil due to the boundary stresses ( 39


contacted surfaces of the specimen to the apparatus were lubricated with silicone
grease to minimize friction in the system.
Whittle et al. (1992) reported the results of a test performed on a dry Ticino
sand reinforced with two-ply steel sheet inclusions. A number of strain gages were
mounted between the two thin steel sheets (0.13 mm thick) to obtain the strain
distribution within the reinforcement. The test results are shown in Figure 2.24. The
figure shows the relationships of the load in the reinforcement versus applied stress
ratio, (R=cri/03). It was concluded that the tensile stress in the reinforcement was a
linear function of the stress ratio, R. It also showed that the maximum tensile stress
occurred at the center of the inclusion and that the tensile stresses in the
reinforcement was minimal when the stress ratio, R, was less than 2.
A plane strain Unit Cell Device (UCD) was developed by Boyle (1995). The
specimen dimensions were 200 mm high, 200 mm wide, and 100 mm deep (/.e., the
longitudinal direction). Figure 2.25 shows schematic diagrams of the apparatus. The
UCD was a load-controlled test apparatus. The vertical pressure was applied by the
top and bottom air bladders to the surfaces of the specimen. The left instrument box
was allowed to move freely in the horizontal direction. The lateral pressure was
applied by the end bladder through the instrument box. The tensile forces at two ends
of the reinforcement layer were measured by load cells. Stiff end plates which were
linked to the clamps controlled that the soil and the reinforcement deform together in
the horizontal direction. The vertical and the horizontal displacements, the major
40


principal stress, and the tension at two ends of the reinforcement layer were measured
directly.
Two different sands, four woven geotextiles, two nonwoven geotextiles, and a
steel sheet were employed in the study. They reported similar results as those of the
previous studies that the reinforcement improved the load carrying capacity of the
dense cohesionless soil as shown in Figure 2.26. The Figure shows the relationships
of the principal stress (ot) versus lateral strain (63). The load carrying capacity of the
reinforced soil specimen reinforced with geotextiles (reinforcing No.l to 6) increased
with the stiffness of the reinforcement that was presented in term of the modulus at
5% strain. The sand reinforced with a steel sheet (No.7) showed significantly higher
deformation-modulus than those with the geotextile reinforcement before yielding
occurred at about 0.3% of lateral strain.
A comparison of the specimen size, soil type, reinforcement types, and
instrumentation of the four plane strain compression tests of reinforced-soil masses
reviewed in this section is presented in Table 2.1. A shortcoming of these triaxial and
plane strain compression tests performed on the GRS mass is their reduced
dimensions. The relatively small dimensions of the test specimens prohibit testing of
a representative reinforced-soil specimen of a typical GRS structure.
41


Figure 220: Behavior of a Unit Ceil With and Without Inclusions: (a) Dense
Sand; (b) Loose Sand (After McGown et aL, 1978)
42


Figure 221: Schematic Diagram of Plane Strain Compression Test Specimen
(After Tatsuoka and Yamauchi, 1936)
43


Z0.0
AVERAGE MINOR FRMCSMt. STRAIN N SOL 2, ()
Figure 2.22: Plane Strain Compression Test Results for Unreinforced and
Reinforced Sand Specimens (After Tatsuoka and Yamauchi, 1986)
44


Figure 2.23: Cross Section Through the APSR Cell (After Larson, 1992)
45


Tfkmm Anlt. **K>,
Figure 224: Stress Distribution in a Steel Inclusion of the AP SR Cell (After
Whittle et al., 1992)
46


Figure 225: Schematic Diagram of the Unit Cell Device: (a) Profile and (b)
Plan View Section (After Boyle, 1995)
47


Figure 226: Unit Cell Device Test Results of Unreinforced and Reinforced
Soils (After Boyle, 1995)
48


Table 2.1: Summary of Four Plane Strain Compression Tests for Reinforced-
SoilMass
Test Dimensions Materials Measurement \
w (mm) H (mm) D (mm) Soil Reinforcement
McGown tt aL (1978) 102 102 152 Leighton Buzzard Sand Aluminium foil Aluminium mesh Non-woven geotextile Vertical Load 1 Vertical deformation 1 Internal deformation 1
Tatsuoka and Yamauchi (1986) 80 75 40 Toyoumsand Brass plates Nonwoven geotextiles Urethane Neoprene Vertical Load Vertical deformation Lateral deformation
Whittle ttaL (1992) 450 570 ISO Ticino sand Steel sheet Vertical Load Vertical deformation Lateral deformation Reinforcement strain
Boyle (1995) 200 200 100 Ottawa sand Gravelly sand Steel sheet Woven gmtextiles Nonwoven geotextiles Vertical Load Vertical deformation Lateral deformation

(AdspladtenTalamkamd Yaauuehi. 1916)
49


3. Laboratory Tests on Solis, Goesynthetics, and
Soil-Geosynthteic Interfaces
Laboratory tests were conducted to examine the behavior of a number of
soils, geosynthetics, and soil-geosynthetic interfaces subject to monotonic loading
and unloading-reloading cycle(s). The laboratory tests consisted of conventional
triaxial compression (CTC) tests for soils, in-isolation load-extension (LE) tests for
geosynthetics, and direct shear (DS) tests for soil-geosynthetic interfaces. Each test
category employed two types of loading sequences: monotonic loading and
unloading-reloading cycle(s). The monotonic-loading tests were conducted to
examine the behavior of the materials and the interfaces subject to monotonic
loading and to provide reference properties for assessing effects of preloading on the
deformation and strength behavior. The unloading-reloading tests were conducted to
examine the behavior subject to unloading-reloading cycle(s) and to assess effects of
preloading on the deformation and strength behavior. Test specimens used for the
monotonic-loading tests were referred to as virgin specimens, whereas test
specimens used for the unloading-reloading tests were referred to as preloaded
specimens.
50


This chapter presents test materials, test descriptions, specimen preparations,
measurement, data reductions, test programs, test results, and discussions of test
results of the laboratory tests.
3.1 Test Materials
3.1.1 Soils
Two types of granular soils were used in this study: an Ottawa Sand and a
Road Base soil, designated as S and RB, respectively. The Ottawa sand was
chosen because of its well-defined properties. The Road Base soil was a granular
material that is commonly used as backfill for GRS retaining walls. It was selected
in this study to examine the behavior of a generic preloaded GRS mass consisting of
a typical construction backfill.
The Ottawa sand used in this study was a subround uniform sand, with its
gradation curve shown in Figure 3.1. The specific gravity of the sand is 2.65. The
maximum and minimum unit weights, per ASTM D854, were 17.65 kN/m3 and
15.34 kN/m3, respectively. The Road Base soil used in this study was a dark brown,
silty sand. It was a backfill material for the preloaded GRS abutments in Black
Hawk, Colorado (Section 2.4.1.4). The soil was classified as SM-SC, per ASTM
D2487. It has 12% of fine particles (passing #200 standard sieve). The gradation
curve is shown in Figure 3.2. The plasticity index and the liquid limit were 6 % and
27 %, respectively. The maximum dry density was 18.75 kN/m3 with the optimum
51


water content of 14.2 %, per ASTM D698. The moisture content-dry unit weight
relationship is shown in Figure 3.3.
3.1.2 Geosynthetics
Two types of geosynthetics, Amoco 2044 and Typar 3301, were used in this
study. Amoco 2044 represents a strong reinforcement material; whereas, Typar 3301
represents a weak reinforcement material.
Amoco 2044 is a woven polypropylene geotextile. The wide-width tensile
strengths, as provided by the manufacturer, in both fill and warp directions are 70
kN/m. Amoco 2044 was a reinforcement material used in the FHWA pier (Section
2.4.1.2) and the Black Hawk abutments (Section 2.4.1.4). Some index properties of
Amoco 2044 are shown in Table 3.1. Typar 3301 is a nonwoven heat-bonded
polypropylene geotextile. It was primarily used for filtration and drainage
applications in actual applications. Some index properties of Typar 3301 are shown
in Table 3.1.
52


9mm
if 1' 9m. Om
0* WMMM
* t I
t t t t t
I
I
i
lllHISSilimiNillllllllllllflllllli
I i m i 1 IH':W i 1 ft JnMML l If
FI c - i : i sf
-1 ; : *: f 8m
i i \ \A !5F jr.-;. - M
*iU ; * , il > 1 rs- .*
u i ! r* s;|; '-'i-. t
i i :;l ir r : - .
U-J #i.> |;l;
Figure 3.1: Grain Size Distribution of Ottawa Sand
53


to i at aoi
Grain Diamatw (mm)
Figure 3 2: Grain Size Distribution of Road Base Soil
Figure 33: Moisture Content-Dry Unit Weight Relationship of Road Base Soil
54


Table 3.1: Some Index Properties of Geosyntehtics
1 Geosynthetic Amoco 2044 Typar3301
Manufacturing method Woven Non-Woven
Wide width tensile strength (ASTM D-4595) 70 kN/m (fill and warp directions) 6 kN/m
Elongation at break (ASTM D-4595) 8% (fill direction) 10% (warp direction) 70%
Grab tensile (ASTM D-4632) 2.22 kN (fill direction) 2.67 kN (warp direction) 0.53 kN
Elongation at break (ASTM D-4632) 20% (fill and warp directions) 60% |
55


3.2 Loading System
The loading system used in this study was the MTS-810 electro-hydraulic
testing system. The MTS-810 testing system comprised a loading frame integrated
with a data acquisition system and a control unit. The loading frame (MTS Model
31131) consisted of four vertical columns that joined a movable crosshead and a
fixed platen (see Figure 3.4). The crosshead was vertically adjustable to
accommodate specimens of various heights. The vertical movement of the crosshead
was controlled by hydraulic crosshead lifts. The crosshead, once in position, locked
into place to prevent slippage during testing. The data acquisition system included a
load cell with a maximum capacity of 1,000 kN (sensitivity = +0.04 kN) and LVDT
(Linear Variable Differential Transformer) with a maximum displacement of 150
mm (sensitivity = +0.03 mm). The LVDT is an electromechanical device that
provides an output voltage which is proportional to the displacement of a moveable
core extension or a stylus. The LVDT was internally mounted on the hydraulic
actuator to provide an indication of the actuator piston rod displacement. The MTS-
810 loading system used the MTS458.20 MicroConsole as a control unit to control
the servohydraulic system.
The data acquisition system and the control unit were connected to an IBM
personal computer. A BASIC software for the control unit and the data acquisition
system developed at the University of Colorado at Denver was modified by the
author for this study. The modified BASIC software provided the inputs (stress-
56


controlled or strain controlled modes) for the control unit and recorded the outputs
(load and displacement) from the data acquisition system. All input parameters must
be predetermined and programmed in the IBM personal computer before starting a
test. For the stress-controlled mode, the failure load of the specimen need to be
estimated beforehand to set an upper bound for the loading magnitude input It is to
be noted that the stress-controlled mode must never be used to load the specimen to
failure. This is because a premature failure may damage the testing system.
3.3 Loading Sequences
In this study, the monotonic loading was applied in a strain-controlled mode
at a constant strain rate. The unloading-reloading cycles were applied in a stress-
controlled mode at a constant loading rate. The unloading-reloading cycles were an
array of different combinations of five loading paths; preloading (PL), unloading to a
zero-load level (UL-Z), unloading to a prestressed load level (UL-PS), reloading
from a zero-load level (RL-Z), and reloading from a prestressed load level (RL-PS).
A diagram illustrating the applied load versus time relationships for I) PL, UL-Z,
and RL-Z paths, and 2) PL, UL-PS, and RL-PS paths are shown in Figure 3.5. In
this report, the term applied load was used to represent different quantities for
different types of tests. It represented the deviator stress in CTC tests, the applied
tensile load in LE tests, the shear stress in DS tests, and the vertical load from the
MTS-810 loading device in SGP tests. The unloading load level (ULL) was equal to
57


the zero-load level for the UL-Z and RL-Z paths, and the prestressed load level
(PSL) for the UL-PS and RL-PS paths. The minimum unloading load level was the
zero-load level. The zero-load level was considered the initial stress of the
specimen; i.e., the applied load = 0.
As shown in Figure 3.5, the preloading (PL) path begins at t = 0 and increases
at a constant loading rate to a preloading load level (PLL). An unloading path (UL-Z
or UL-PS path) follows the PL path. The UL-Z path involves a decrease of the
loading magnitude from a preloading load level to a zero-load level {i.e., ULL = 0).
The RL-Z path follows the UL-Z path. The UL-PS path involves a decrease of the
loading magnitude from a preloading load level to a prestressed-load level {i.e., ULL
= PSL). The RL-PS path follows the UL-PS path. The preloading path resumes
when the magnitude of the load in the reloading path exceeds the preloading load
level.
58


FLOOR-MOUNTED LOAD FRAME
Figure 3.4: MTS-810 Loading System
59


Applied Load
PLL
PL/ \uL
ULL |___
*
RL
Time
Two unloading-reloading paths
1) When ULL = 0 => UL-Z, and RL-Z paths
2) When ULL = PSL => UL-PS, and RL-PS paths
Figure 3.5: General Loading Sequences
60


3.4 Conventional Triaxial Compression Tests for Soils
A series of CTC tests under monotonic loading and unloading-reloading
cycles were conducted on the two soils described in Section 3.1.1. The purposes of
the tests were to examine the behavior of the soils subject to unloading-reloading
cycles and to calibrate soil model parameters in the finite element analysis and the
SPR model (Sections 5.3 and 6.3).
3.4.1 Test Description
The CTC test was performed on an unsaturated soil specimen. The triaxial
chamber was place in the MTS loading device as shown in Figure 3.6. Confining
pressure was applied to the specimen by pressurizing the water surrounding the
specimen. The applied vertical load and the vertical displacement of the loading rod
were recorded by a data acquisition system integrated with the MTS-810 loading
device. The volume-change occurred during shear was measured by monitoring the
volume change of water entering or leaving the triaxial chamber.
61


Figure 3.6: Conventional Triaxial Compression Test Apparatus
62


3.4.2 Specimens Preparation and Test Procedure
The Ottawa sand specimen in the triaxial tests was 158 mm in height and 71
mm in diameter. The specimen was prepared at a unit weight of 16.85 kN/m3 (+0.15
kN/m3). The specimen preparation and the test procedure for the Ottawa sand are
described as follows:
1. Obtain the dry Ottawa sand from the batch;
2. Use o-rings to attach a 0.2-mm thick rubber membrane to the base platen;
3. Place a porous stone at the base of the paten;
4. Place a metallic mold (a split-barrel type) around the rubber membrane and
fold the top portion of the membrane down and over the mold;
5. Pour the sand in the mold by using 50-mm diameter funnel with the opening
diameter of 5 mm at a constant drop height of approximately 80 mm;
6. Place a porous stone on top of the specimen;
7. Place the top platen on the porous stone and roll the rubber membrane over
the top platen and seal to the circumference of the top platen with o-rings;
8. Apply a vacuum pressure of 35 kPa on the specimen through the back
pressure valve which was connected to the base of the specimen;
9. Remove the metallic mold and obtain the average height and average
diameter of the specimen by using a stand ruler and a 7c-tape;
10. Place a lucite cylinder on the cell base;
11. Place the triaxial chamber in a MTS-810 loading frame;
63


12. Apply a predetermined confining pressure and open the back pressure valve;
13. After fifteen minutes of consolidation with the confining pressure, start to
apply shear stress at a prescribed strain or loading rate:
14. Record the vertical applied load, the axial displacement, and the volume
change of the specimen during shear.
The Road Base soil specimen in the CTC tests was 305 mm in height and 152
mm in diameter. The specimen was prepared at a dry unit weight of 17.81 kN/m3
(+0.1 kN/m3) with a water content of 12.2 %. The specimen preparation and the test
procedure for the Road Base soil are described as follows:
1. Prepare the soil at the desired moisture content of 12.2% and cure overnight
in a sealed container inside a high humidity room;
2. Use o-rings to attach a 0.3-mm thick rubber membrane to the base platen;
3. Place a porous stone at the base of the platen;
4. Place a metallic mold (a split-barrel type) around the rubber membrane and
fold the top portion of the membrane down and over the mold;
5. Compact the soil inside the mold in twelve layers by a 4-lb standard Proctor
hammer at the prescribed density;
6. Place a porous stone on top of the specimen;
7. Place the top platen on the porous stone and roll the rubber membrane over
the top platen and seal to the circumference of the top platen with o-rings;
64


8. Apply a vacuum pressure of 35 kPa on the specimen through the back
pressure valve which was connected to the base of the specimen;
9. Remove the metallic mold and attach a second layer of rubber membrane to
the specimen with o-rings on the top and base platens;
10. Obtain the average height and average diameter of the specimen by using a
stand ruler and a tt-tape;
11. Place a Iucite cylinder on the cell base;
12. Fill the triaxial chamber with water;
13. Place the triaxial chamber in a MTS-810 loading frame;
14. Apply a predetermined confining pressure and open the back pressure valve;
15. After one hour of consolidation with the confining pressure, start to apply
shear stress at a prescribed strain or loading rate;
16. Record the vertical applied load, the axial displacement, and the volume
change of the specimen during shear.
3.43 Measurement and Test Data Reduction
The change in height of the specimen and the net applied axial load was
measured by the data acquisition system of the MTS-810 testing system. The axial
strain (8a) and the deviator stress (01-03) were calculated by the following formulas:

AH
(3-1)
65


A-..........(3-2)
Ar
corr
4
(1-0
(3.3)
where
AH = change in height of the specimen
Hq = specimen height after consolidation
P = net applied axial load
Acor. = corrected area of the specimen during shear
Ao = area of the specimen after consolidation
The volume change of the specimen was measured by monitoring the volume
of water entering or leaving the triaxial cell to compensate for the change in volume
of the specimen. A Validyne transducer (model DPI5-30) was used to monitor the
change of water level in a burette that was connected to the surrounding water in the
triaxial cell. The Validyne transducer was connected to an IBM personal computer.
The sensitivity of the transducer was 0.01 cm3. Corrections of the measured values
from the transducer were made to account for expansions of the triaxial chamber and
the tube and penetration of the loading ram into the triaxial chamber during shear.
The volumetric strain (ev) was calculated by the following formula:
66


S.
(3.4)
AV
r0
where AV = volume change of the specimen
V0 = volume of specimen after consolidation
3.4.4 Test Programs
Test programs for the CTC tests are presented in Table 3.2 for the Ottawa
sand and Table 33 for the Road Base soil. The test program was divided into two
groups: monotonic-loading (M) tests and unloading-reloading (UR) tests.
Designations of all the tests are shown in the Tables. The monotonic-loading tests
were conducted in a strain-controlled mode at a constant strain rate of 0.5 % per
minute. The unloading-reloading tests were conducted in a stress-controlled mode
with various loading sequences at a constant loading rate of 10 kfa per minute and
followed by a strain-controlled mode at a constant strain rate of 0.5 % per minute
until failure occurred.
67


Table 32: CTC Test Program for the Ottawa Sand
Tot
Designation
Confining
Front*
Loading Sequence
T-M-Sl
T-M-S2
T-M-S3
(fcPe)
69
207
MS
Strain-Controlled
t-m-mastom
T-UR-St
Stross-Controilod
Strain-Con trotted
r-WUOtTest
T-UR-ai Teat
T-UR-S2
207
Strona-ControUod
Strain-Controlled
r-UUXTeat
T-UR-S3
MS
Straae-ControBod
Strain-Controlled
T-tM-MTeet
T-UR-OlTeM
0 500 1000
Time (min.)
68


Table 33: CTC Test Program for the Road Base Soil
Tot
Dcrifurioa
Cuuiiof
Pramrc
Loading Sequence
T-M-RB1
T-M-RB2
T-M-RB3
(kfe)
69
207
MS
Strain-Controlled
TJMtruSTeata
T-UR-RBI
69
Slraaa-Con IraU ad
ruuuiTm
Strain-Controlled
T-UttftBITaat
Tlan (aunj
T-UR-RB2
207
Straae-ControRed
T-uuunrm
Strain-Controlled
T-mtauieet
T-UR.RB3
MS
Stnaa-ContraOad
Strain-Controlled
raaMtasTeoc
T-UIUtaiTeM
TIhm (min.)
69


3.4.5 Test Results and Discussions
The CTC test results and discussions of the test results are presented in the
following sections. The general behavior was first described, followed by an
assessment of the effects of preloading on deformation and shear strength of the
soils.
3.4.5.1 General Behavior
The peak or maximum deviator stress generally corresponds to the failure
state of the specimen. When the deviator stress continued to increase without
showing a peak value, the deviator stress at 10 % axial strain was considered the
maximum deviator stress. A positive sign of the volumetric strain represents
specimen dilation; whereas, a negative sign represents specimen contraction.
Figures 3.7 and 3.8 show the results of the monotonic-loading CTC tests.
The deviator stress increased with the axial strain until failure occurred. The Ottawa
sand initially contracted during shear and started to dilate at axial strains less than 0.8
%. Similar to the sand, the Road Base soil at lower confining pressures contracted
initially and dilated after it reached certain axial strains. The Road Base soil
experienced dilatancy at larger axial strains (1.2 % and 6 % at confining pressures of
69 kPa and 207 kPa, respectively) than the Ottawa sand. At the confining pressure
of345 kPa, the Road Base soil specimen did not show the dilatant behavior.
Figures 3.9 to 3.14 show the results of the unloading-reloading CTC tests.
The stress-strain curve was similar to the typical stress-strain curve of a soil
70


specimen in the CTC test subject to primary loading, unloading, and reloading as
described in Section 2.1.
The specimen was initially loaded to a preloading load level then unloaded.
Irrecoverable and recoverable deformations occurred when the specimen was
unloaded. This behavior may be explained in terms of deformation of grains and
sliding between grains (see, e.g., Makhlouf and Stewart, 1965; Ko and Scott, 1967;
Lade and Duncan, 1976). The recoverable deformation was due to the elastic
deformation of individual grains, whereas the irrecoverable deformation was
primarily caused by the sliding between individual grains. In the primary loading,
both the elastic deformation and sliding between grains occurred. Upon unloading,
individual grains did not rebound to their original positions but remained
approximately in their displaced positions. This behavior caused the irrecoverable
deformation.
During the initial unloading path in which the deviator stress started to
reduce, it was observed that axial and volumetric strains still continued to behave in
the same fashion as those in the PL path. Specifically, the downward deformation
continued and the volume change behavior was in expansion for the Ottawa sand and
in contraction for the Road Base soil. This behavior has been reported in cyclic
triaxial tests as a rounded comer of hysteresis loops (Hyodo et al., 1994) and
explained in terms of creep of a soil specimen by Tatsuoka and Shibuya (1991). The
71


magnitude of the creep deformation was found to be more significant when the
specimen was unloaded from a high preloading load level.
It is the authors opinion that the influence of soil creep at an initial
unloading path is suppressed in an unloading-reloading CTC test conducted in a
strain-controlled mode. This is believed to be a strain-controlled loading
characteristic. Under the strain-controlled mode, the vertical deformation of the
specimen is controlled by vertical movements of the loading rod. It is tacitly
assumed that the specimen is in full contact with the loading rod. The unloading
path begins when the loading rod movement is reversed. This unloading mechanism
eliminates the possibility of the soil to continue the vertical downward deformation
during the unloading path.
When the specimen was unloaded and reloaded from small to moderate
preloading load levels, it behaved approximately linear-elastically. However, when
the unloading path took place at a high preloading load level (i.e., close to failure),
the linearity between stress and strain may not be assumed as a result of significant
hysteresis loops. The hysteresis loop existed in all unloading-reloading portions of
deviator stress-axial strain curves. The width of the hysteresis loop is typically
larger at the high preloading load level.
72


Volumetric Strain (%) Deviator Stress (kPa)
Figure 3.7: Test Results of Monotonic-Loading CTC Tests on Ottawa Sand
(Tests T-M-SI, 2, and 3)
73


Figure 3.8: Test Results of Monotonic-Loading CTC Tests on Road Base
Soil
(Tests T-M-RBI,2, and 3)
74


.9: Results of Test T-UR-S1 (Confining Pressure
Volumetric Strain (%)
Deviator Stress (kPa)
o 8 8 8 8 8


Volumetric Strain (%) Deviator Stress (kPa)
Axial Strain (%)
Figure 3.10: Results of Test T-UR-S2 (Confining Pressure = 207 kPa)
76


Volumetric Strain (%) Deviator Stress (kPa)
Figure3.Il: Results of Test T-UR-S3 (Confining Pressure = 345 kPa)
77


Deviator Stress (kPa)
O 8 8 S 8 8 8
O


s
Axial Sfran(%)
Deviator Stress (kPa)


Volumetric Strain (%) Deviator Stress (kPa)
Figure 3.14: Results of Test T-UR-RB3 (Confining Pressure = 345 kPa)
80


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