Citation
Deterioration studies as part of bridge infrastructure management program for the city and county of Denver

Material Information

Title:
Deterioration studies as part of bridge infrastructure management program for the city and county of Denver
Creator:
Guenther, Daniel James
Publication Date:
Language:
English
Physical Description:
xiii, 162 leaves : illustrations ; 28 cm

Thesis/Dissertation Information

Degree:
Master's ( Master of science)
Degree Grantor:
University of Colorado Denver
Degree Divisions:
Department of Civil Engineering, CU Denver
Degree Disciplines:
Civil engineering

Subjects

Subjects / Keywords:
Bridges -- Maintenance and repair -- Colorado -- Denver ( lcsh )
Bridges -- Deterioration -- Colorado -- Denver ( lcsh )
Infrastructure (Economics) -- Management -- Colorado -- Denver ( lcsh )
Bridges -- Deterioration ( fast )
Bridges -- Maintenance and repair ( fast )
Infrastructure (Economics) -- Management ( fast )
Colorado -- Denver ( fast )
Genre:
bibliography ( marcgt )
theses ( marcgt )
non-fiction ( marcgt )

Notes

Bibliography:
Includes bibliographical references (leaves 160-162).
General Note:
Department of Civil Engineering
Statement of Responsibility:
by Daniel James Guenther.

Record Information

Source Institution:
|University of Colorado Denver
Holding Location:
Auraria Library
Rights Management:
All applicable rights reserved by the source institution and holding location.
Resource Identifier:
435778636 ( OCLC )
ocn435778636
Classification:
LD1193.E53 2009m G83 ( lcc )

Full Text
DETERIORATION STUDIES AS PART OF BRIDGE INFRASTRUCTURE MANAGEMENT PROGRAM
FOR THE CITY AND COUNTY OF DENVER
By
Daniel James Guenther B.S. University of Colorado Denver, 2007
A thesis submitted to the University of Colorado Denver In partial fulfillment of Master of Science Civil Engineering 2009
i


This thesis for the Master of Science
Civil Engineering degree by
Daniel James Guenther has been approved by
Kevin L. Rens
Stephan Durham
Bruce Janson
04-1 SO) Q-$


Guenther, Daniel J. (M.S., Civil Engineering)
Studies of Bridge Deterioration as Part of Infrastructure Management Program for The City and County of Denver
Thesis directed by Professor Kevin L. Rens, Ph.D., P.E.
ABSTRACT
Four bridge structures were evaluated to determine structural interaction and/or the cause and extent of distress observed during routine inspections. The first investigation included finite element method analysis to determine differential interactions between the three independent structures that make up the Quebec over Airlawn Bridge. Longitudinal joints between the structures had failed and subsequent leakage had led to the deterioration of the deck and adjacent superstructure elements. Structural analysis was warranted to determine the feasibility of monolithically joining the decks due to inconsistencies between adjacent superstructure designs. Differential deflections and rotations were calculated to aid in the feasibility analysis. The second investigation included non-destructive testing and a chemical analysis in an effort to determine the origin and effects of leakage adjacent to a post-tension duct within the 20th Street HOV Viaduct. Visual inspection, sounding, ultrasonic velocity testing including tomography and rebound hammer testing were completed to investigate the material properties and voids within the box girder web. Chemical analysis on the efflorescence was used to determine the origin of the leakage. Non-destructive analysis indicated no unwanted voids within the duct or within the web of


the member and the chemical content of the efflorescence indicated that the likely source
of the water was through cracks in the overlying deck. The third investigation included a structural analysis of the abutments and superstructures of the City of Cuernavaca Park Pedestrian Bridges. Despite previous repairs, cracking of the backwall of the abutments in the park were re-manifesting. FEM models determined that the stresses were due to thermal loading of the abutment and its interaction with the superstructure. Analysis revealed the cracking was likely due to positive thermal loading and non-functioning moveable bearings. The fourth investigation included non-destructive testing of the cap beam of bent #6 of the Evans over Santa Fe Bridge. Severe delamination, cracking and a large spall under a bearing warranted investigation into the extent of damages to determine structural adequacy and repair options. The extent of damages was found to be limited to the extent of previous repairs on the beam with the exception of an extreme propagation of a large crack.
This abstract accurately represents the content of the candidate's thesis. I recommend its
publication.
Signed
Kevin L. Rens


DEDICATION
This thesis is dedicated to my wife, my father, and my mother. Without the support and persistence of these, I would not have nor could I have taken this step.


ACKNOWLEDGEMENT
I would like to acknowledge and thank Kevin Rens who has given me many opportunities and valued advice throughout my studies and beyond. I would also like to acknowledge the City and County of Denver, especially the infrastructure management personnel, who have provided me with valuable opportunities both academically and financially


TABLE OF CONTENTS
Figures.............................................................x
Tables..............................................................xii
Chapter
1. PROGRAM OVERVIEW..................................................14
1.1 INTRODUCTION..................................................14
1.2 GIS...........................................................15
1.3 ALLEYS........................................................16
1.4 CURB AND GUTTER...............................................16
1.5 PAVEMENT......................................................17
1.6 FLY ASH.......................................................17
1.7 BRIDGE MANAGEMENT.............................................18
1.8 CONCLUSION....................................................19
2. INVESTIGATION TECHNIQUES..........................................21
2.1 INTRODUCTION..................................................21
2.2 VISUAL INSPECTION.............................................21
2.3 SOUNDING......................................................23
2.4 ULTRASONIC VELOCITY...........................................23
2.5 REBOUND HAMMER................................................24
2.6 TOMOGRAPHY....................................................25
2.7 CONCLUSION....................................................25
3. QUEBEC OVER AIRLAWN BRIDGE........................................26
3.1 OVERVIEW......................................................26
3.2 LOCATION......................................................26
3.3 DESCRIPTION...................................................26
3.4 HISTORY.......................................................28
3.5 PROBLEM.......................................................30
3.6 ANALYSIS......................................................32
3.7 RESULTS.......................................................36
vii


3.8 REPAIRS.........................................................37
3.9 CONCLUSION......................................................38
4. 20 STREET HOV VIADUCT..............................................39
4.1 OVERVIEW........................................................39
4.2 LOCATION........................................................39
4.3 DESCRIPTION.....................................................41
4.4 HISTORY.........................................................42
4.5 PROBLEM.........................................................43
4.6 ANALYSIS........................................................45
4.6.1 SETUP.......................................................45
4.6.2 VISUAL INSPECTION...........................................46
4.6.3 REBOUND HAMMER..............................................48
4.6.4 ULTRASONIC VELOCITY TESTING.................................50
4.6.5 CHEMICAL ANALYSIS OF EFFLORESCENCE..........................53
4.7 CONCLUSION......................................................55
5. CITY OF CUERNAVACA PARK PEDESTRIAN BRIDGES..........................57
5.1 OVERVIEW........................................................57
5.2 LOCATION........................................................57
5.3 DESCRIPTION.....................................................58
5.4 HISTORY.........................................................60
5.5 PROBLEM.........................................................60
5.6 VISUAL INSPECTION...............................................61
5.7 ANALYSIS........................................................65
5.7.1 MODEL 1 BRIDGE AND DECK TRUSS.............................66
5.7.2 MODEL 2: CONCRETE BRIDGE ABUTMENT...........................69
5.8 Model 3 H-Pile................................................71
5.9 RESULTS.........................................................74
5.10 CONCLUSION......................................................84
5.11 REPAIRS.........................................................85
6. EVANS OVER SANTA FE BRIDGE..........................................86
6.1 OVERVIEW........................................................86
6.2 DESCRIPTION.....................................................86
6.3 HISTORY.........................................................88
6.4 PROBLEM.........................................................90
viii


6.5 ANALYSIS AND RESULTS................................................90
6.5.1 VISUAL INSPECTION...............................................91
6.5.2 SOUNDING........................................................92
6.5.3 REBOUND HAMMER..................................................93
6.5.4 ULTRASONIC VELOCITY INCLUDING TOMOGRAPHY........................93
6.6 CONCLUSION..........................................................95
7. SUMMARY AND CONCLUSIONS..................................................98
A. APPENDIX A CRACK MAPPING OF 8th AVENUE VIADUCT PIERS.................100
A.l PIER 2.............................................................100
A.2 PIER 3.............................................................102
A.3 PIER 4.............................................................104
A.4 PIER 5.............................................................106
A.5 PIER 6.............................................................108
A.6 PIER 7.............................................................110
A.7 PIER 8.............................................................112
A.8 PIER 9.............................................................114
A.9 PIER 10............................................................116
A.10 PIER 11............................................................118
A.11 PIER 12............................................................120
A.12 PIER 13............................................................122
A.13 PIER 14............................................................124
A.14 PIER 15............................................................126
A.15 PIER 16............................................................128
A. 16 PIER 17............................................................130
B. APPENDIX B Photographic Log of Studies................................132
B. l QUEBEC OVER AIRLAWN................................................133
B.2 20th Street HOV Viaduct............................................140
B.3 City of Cuernavaca Park Pedestrian Bridges.........................149
B.4 Evans Over Santa Fe................................................156
BIBLIOGRAPHY................................................................160
IX


LIST OF FIGURES
Figure
3-1 VIEW OF SUBSTRUCTURE AND SUPERSTRUCTURE...........................27
3-2 SPALLING AND EFFLORESCENCE ALONG LONGITUDINAL JOINT...............30
3-3 SUBSTRUCTURE PLAN VIEW............................................32
3-4 SUPERSTRUCTURE SHELL ELEMENT MODEL................................33
3-5 COMBINED SUPERSTRUCTURE ELEMENT MODEL.............................33
3- 1 CRITICAL STRESS AREAS............................................35
4- 1 VIEW OF SUPERSTRUCTURE AND SUBSTRUCTURE OF 20TH STREET HOV VIADUCT.40
4-2 LOCATION AND ORIENTATION OF 20TH STREET VIADUCT...................40
4-3 EFFLORESCENCE AT HORIZONTAL CRACK.................................44
4-4 GRID LAYOUT, BOLD LINE IS HORIZONTAL CRACK........................46
4-5 INTERIOR FACE OF THE NORTH WEB....................................47
4-6 EXTERIOR FACE OF THE NORTH WEB....................................48
4-7 INTERIOR FACE REBOUND HAMMER RESULTS TOPOGRAPHIC..................49
4-8 EXTERIOR FACE REBOUND HAMMER RESULTS TOPOGRAPHIC..................49
4-9 INTERIOR FACE REBOUND HAMMER RESULTS SHADED.......................50
4-10 EXTERIOR FACE REBOUND HAMMER RESULTS SHADED........................50
4-12 PLAN VIEW OF ULTRASONIC VELOCITY RAY PATHS.........................51
4-11 SIDE VIEW OF ULTRASONIC VELOCITY RAY PATHS.........................51
4-13 DIRECT METHOD CONCRETE QUALITY.....................................52
4-14 TOMOGRAPHIC SECTION CUT THROUGH LOWER PRESTRESS DUCT...............53
4- 15 CHEMICAL ANALYSIS RESULTS 20TH STREET HOV VIADUCT.................54
5- 1 OVERVIEW OF CITY OF CUERNAVACA PARK..............................57
5-2 ELLIPTICAL LAYOUT OF BRIDGES AND FOOT PATH........................58
5-3 PHOTO OF PEDESTRIAN BRIDGE ABUTMENT...............................59
5-4 MOVEABLE BEARING ASSEMBLY.........................................62
5-5 TYPICAL FIXED BEARING ASSEMBLY....................................62
5-6 MISSING GUIDE BOLT IN GUIDE ASSEMBLY................................63
5-7 TYPICAL VERTICAL CRACK IN ABUTMENT BACKWALL.........................63
5-8 TYPICAL CRACKING AT ABUTMENT CORNER.................................64
5-9 LOCATION OF CRACKS ON BACK WALL.....................................65
5-10 SUPERSTRUCTURE AND DECK MODEL....................................68
5-11 BRIDGE ABUTMENT MODEL............................................70
5-12 ABUTMENT MODEL INCLUDING STIFFNESS OF H-PILE FOUNDATION............73
5-13 ABUTMENT STRESSES UNDER POSITIVE THERMAL LOADING...................75
5-14 ABUTMENT STRESSES UNDER NEGATIVE THERMAL LOADING...................76
5-15 POSITVE THERMAL LOADING SUPERIMPOSED WITH PIN-PIN REACTIONS........78
5-16 NEGATIVE THERMAL LOADING SUPERIMPOSED WITH PIN-PIN REACTIONS.......79
5-17 POSITIVE THERMAL LOADING SUPERIMPOSED WITH PIN-SPRING REACTIONS....80
x


5-18 NEGATIVE THERMAL LOADING SUPERIMPOSED WITH PIN-SPRING REACTIONS..81
5-19 POSITIVE THERMAL LOADING SUPERIMPOSED WITH PIN-ROLLER REACTION...82
5- 20 NEGATIVE THERMAL LOADING SUPERIMPOSED WITH PIN-ROLLER REACTIONS.83
6- 1 PLAN VIEW OF EVANS OVER SANTA FE BRIDGE..........................87
6-2 PLANS FOR REPAIR OF DELAMINATED AREAS.............................89
6-3 EAST FACE OF SOUTH PIER CAP.......................................91
6-4 WEST FACE OF SOUTH PIER CAP.......................................91
6-5 DELAMINATION ON EAST FACE OF BENT #6..............................92
6-6 DELAMINATION ON WEST FACE OF STRUCTURE............................93
6-7 TOMOGRAPHIC REPRESENTATION OF THE NORTH END CAP...................94
6-8 TOMOGRAPHIC REPRESENTATION OF THE SOUTH END CAP...................94
6-9 PREVIOUS REPAIR COMPARED TO EXISTING DELAMINATION OF WEST FACE....96
6-10 PREVIOUS REPAIR COMPARED TO EXISTING DELAMINATION OF EAST FACE...96
A-l PIER 2 CRACK MAP.................................................100
A-2 PIER 3 CRACK MAP.................................................102
A-3 PIER 4 CRACK MAP.................................................104
A-4 PIER 5 CRACK MAP.................................................106
A-5 PIER 6 CRACK MAP.................................................108
A-6 PIER 7 CRACK MAP.................................................110
A-7 PIER 8 CRACK MAP.................................................112
A-8 PIER 9 CRACK MAP.................................................114
A-9 PIER 10 CRACK MAP................................................116
A-10 PIER 11 CRACK MAP...............................................118
A-ll PIER 12 CRACK MAP...............................................120
A-12 PIER 13 CRACK MAP...............................................122
A-13 PIER 14 CRACK MAP...............................................124
A-14 PIER 15 CRACK MAP...............................................126
A-15 PIER 16 CRACK MAP...............................................128
A-16 PIER 17 CRACK MAP...............................................130
B-l LONGITUDINAL OF EAST STRUCTURE....................................133
B-2 HINGE AT CENTER STRUCTURE.........................................133
B-3 WEST AND CENTER STRUCTURE LOOKING NORTH...........................134
B-4 CENTER AND EAST STRUCTURE.........................................135
B-5 OVERALL OF CENTER STRUCTURE BENT..................................135
B-6 OVERALL OF NORTH BENT.............................................136
B-7 CENTER SPAN AND DIAPHRAGMS........................................136
B-8 DETERIORATION AT WEST LONGITUDINAL JOINT.........................137
B-9 NORTH BEARING....................................................137
B-10 NORTH BENT BEARING..............................................138
B-ll OVERALL OF SOUTH BENT...........................................138
B-12 CLOSE-UP OF LONGITUDINAL CRACK..................................140
B-13 INTERIOR INSPECTION AREA........................................141
XI


B-14 FORMED SLOT IN WALL...............................................141
B-15 INTERIOR RIGHT OF INSPECTION AREA.................................142
B-16 INTERIOR CENTER RIGHT OF INSPECTION AREA..........................142
B-17 INTERIOR CENTER OF INSPECITON AREA................................143
B-18 INTERIOR LEFT OF INSPECTION AREA..................................143
B-19 EXTERIOR LEFT OF INSPECTION AREA..................................144
B-20 EXTERIOR CENTER OF INSPECTION AREA................................144
B-21 EXTERIOR CENTER LEFT OF INSPECTION AREA...........................145
B-22 EXTERIOR LEFT OF INSPECITON AREA..................................145
B-23 OVERALL OF EXTERIOR INSPECITON AREA...............................146
B-24 INSPECTION AREA FROM GROUND LEVEL.................................146
B-25 HINGE NEAR INSPECTION AREA........................................147
B-26 OVERALL FROM GROUND LEVEL.........................................147
B-27 LONGITUDINAL VIEW OF SPAN.........................................148
B-28 LONGITUDINAL OF BRIDGE LOOKING WEST...............................148
B-29 DETERIORATION AT NORTHEAST ABUTMENT...............................149
B-30 DETERIORATION AT NORTHEAST ABUTMENT...............................150
B-31 LATERAL DEFLECTION AT NORTHWEST ABUTMENT..........................150
B-32 DETERIORATION OF SOUTHEAST ABUTMENT...............................151
B-33 FIXED BEARING AT SOUTHEAST ABUTMENT...............................151
B-34 EXPOSED REBAR AT NORTHWEST ABUTMENT...............................152
B-35 EXPOSED REABER AT NORTHWEST ABUTMENT..............................152
B 36 MOVEABLE BEARING AT PIER..........................................153
B-37 BOTTOM CHORD OF NORTH STRUCTURE...................................153
B-38 VERTICAL CRACK AT SOUTHWEST ABUTMENT..............................154
B-39 MOVEABLE BEARING AT PIER..........................................155
B-40 EXPANSION JOINT AT SOUTHWEST ABUTMENT.............................155
B-41 WEST FACE OF PIER CAP.............................................156
B-42 EAST FACE OF PIER CAP INCLUDING PREVIOUS REPAIR...................157
B-43 EAST FACE OF PIER CAP.............................................157
B-44 EAST FACE OF SOUTH PIER CAP.......................................158
B-45 NORTH FACE OF SOUTH PIER CAP......................................158
B-46 BOTTOM FACE OF SOUTH PIER CAP.....................................159
B-47 SOUTH FACE OF SOUTH PIER CAP......................................159
xii


LIST OF TABLES
Table
2- 1 RELATIONSHIP BETWEEN CONCRETE QUALITY AND VELOCITY (Rens & Taewan, 2007). 24
3- 0-1 ANALYSIS RESULTS OF DIFFERENTIAL MOVEMENT OF FRAME AND SHELL MODEL.36
3-0-2 ANALYSIS RESULTS DIFFERENTIAL MOVEMENTS OF SHELL MODEL..............36
5-1 ENVELOPE REACTIONS BASED ON BEARING ASSUMPTIONS.......................77
A-l PIER 2 CRACK LENGTH TABLE..........................................101
A-2 PIER 3 CRACK LENGTH TABLE..........................................103
A-3 PIER 4 CRACK LENGTH TABLE..........................................105
A-4 PIER 5 CRACK LENGTH TABLE..........................................107
A-5 PIER 6 CRACK LENGTH TABLE..........................................109
A-6 PIER 7 CRACK LENGTH TABLE..........................................Ill
A-7 PIER 8 CRACK LENGTH TABLE..........................................113
A-8 PIER 9 CRACK LENGTH TABLE..........................................115
A-9 PIER 10 CRACK LENGTH TABLE............................................117
A-10 PIER 11 CRACK LENGTH TABLE........................................119
A-ll PIER 12 CRACK LENGTH TABLE........................................121
A-12 PIER 13 CRACK LENGTH TABLE........................................123
A-13 PIER 14 CRACK LENGTH TABLE........................................125
A-14 PIER 15 CRACK LENGTH TABLE........................................127
A-15 PIER 16 CRACK LENGTH TABLE........................................129
A-16 PIER 17 CRACK LENGTH TABLE........................................131
xili


1. PROGRAM OVERVIEW
1.1 INTRODUCTION
Since 1997, The University of Colorado Denver (UCD) and the City and County of Denver (CCD) have worked together in a cooperative research program to provide engineering opportunities to students and to the City. The program has worked with a variety of City departments including administration, engineering, geographical information systems (GIS), transportation, and Infrastructure Planning and Programming (IPP) among others. The students have received valuable work experience, networking opportunities, job placement opportunities, research projects, and the opportunities to work with professionals under the supervision and tutelage of academic faculty.
The program is structured such that the student scheduling, payroll, and benefits are handled by the University while the tasks, work detail, and assignments are provided by the City. Supervision is jointly handled by CCD and UCD personnel. This cooperative environment is mutually beneficial. The relationship has provided the University with many research and academic studies that may have otherwise been contracted to consultants or been disregarded for lack of resources. The program also offers the University the ability to offer financial aid in the form of employment, which has attracted many students from local regions and from abroad. The City benefits from the academic scrutiny and faculty review
14


of work assignments without the expense usually associated with University or higher level
involvement. The interns also provide assistance with clerical and other typical office tasks.
As of February 2009, 86 interns had been employed through the research program. There had been as many as 11 concurrent projects running at one time and as many as 20 interns employed at once. The program continues to grow. The following is a summary of a sample of City projects in which the University has played an integral role.
1.2 GIS
The GIS department has routinely requested interns for assistance with layer maintenance. Students have performed visual verification, database filtering, data input, and assisted with layer construction. The University currently offers a GIS degree program and several students have gained valuable experience and knowledge related to their academic pursuits while working with the City GIS department. In addition, all infrastructure data are managed by GIS.
Rens, K. L., Nogueira, C.L., Neiman, Y., Gruber, T., and Johnson, L. (1999). Bridge management systems for the City and County of Denver. Pract. Period. Struct. Des. Constr., 4(4). 131-136.
Rens, K. L., Neiman, Y., Nogueria, C. (1999). Application of a Geographic Information System for Bridge Management, Deterioration Modeling, and Condition Assessment. Structural Faults and Repair. London, England. 131-136.
15


1.3 ALLEYS
The IPP department has utilized the largest number of students. The department is tasked with maintenance of the City's infrastructure. Included among the maintenance items are the alleys that are part of the public right-of-way. The research program has provided routine inspectors to inventory and asses the condition of all 6,002 alleys in the City. The City uses the information collected by the students to prioritize maintenance tasks. Once all the alleys have been inspected, the process starts over again, allowing for a continual condition assessment of the alleys. The University has assisted with the development of the organization, quantification, and rating process now used by the City. The following research papers have been developed as a result of the Alleys project:
Chavooshi, S. Analysis of alley rigid pavements in the City and County of Denver, Colorado. Denver, CO: Univ. of Colorado at Denver, 2006.
Sathantip, C. Alley management system for the City and County of Denver. Denver, CO: Univ. of Colorado at Denver, 2002.
1.4 CURB AND GUTTER
Very similar to the alleys project, the research program has assisted with condition assessment of the City's curb and gutters. This project works in conjunction with the GIS department to locate and label distresses to the system. In the past, students were outfitted with a voice-command global positioning unit which was used to pinpoint and map distresses. The program has evolved to incorporate a handheld GIS device that allows the students to work directly with the layers to streamline the process. The program is
16


currently in its second cycle of assessing the 3,300 miles of curb and gutter maintained by
the city. The following papers were written with regards to this project.
Rens, K. L. (2007). Inventory and assessment of Denver, Colorado: curb and gutters. Journal of Performance of Constructed Facilities (May/June), 249-254.
Usagani, R. City and County of Denver, Colorado, curb and gutter management system. Denver, CO: Univ. of Colorado at Denver, 2004.
1.5 PAVEMENT
Like the alleys and curb and gutter programs, inspection of City pavement is comleted by the interns. Street overlays are inspected for spalls, cracks, pot holes, and other distresses. The students input quantities of distress assigned to individual street segments. An algorithm that outputs a condition index is then applied to the data, which is then used to prioritize overlay maintenance. Currently ,the students are just completed the second cycle of pavement inspection of the 1,600 miles of pavement in Denver.
Rens, K. L., and Staley, B. (2009). Rating City and County of Denver Urban Pavement Constructability Using a Profiler. Pract. Period. Struct. Des. Constr.
1.6 FLY ASH
The fly ash project is a good example of the mutually beneficial relationship the research program provides. The University was researching the structural and durability aspects of concrete with high fly ash content. The City was looking for ways to be more economical and more sustainable with infrastructure construction and repair projects. Working in
17


concert with the needs of the City, the University provided concrete mix designs that were
incorporated into infrastructure projects and closely monitored for performance. Both organizations are hopeful that the performance of the designs provides a basis for more sustainable concrete design specifications. The following paper discusses the application and results of the project.
Reiner, M., and Rens, K. (2006). High volume fly ash concrete: Analysis and application. Pract. Period. Struct. Des. Constr., 11(1), 58-64.
Reiner, M., Rens, K. L., and Ramaswami, A. The Role of HVFA concrete in the Sustainability of the Urban Build Environment. Journal of Engineering for Sustainable Development., 1(1). 43-54.
1.7 BRIDGE MANAGEMENT
Bridge infrastructure in Denver, as in other parts of the country, has begun to surpass its design life expectancy. As such, maintenance of the structures involves in-depth analysis to determine the cause and extent of distresses. The University has been able to offer nondestructive and destructive analysis to aid in the maintenance of the bridge infrastructure. The following papers have been written regarding the analysis of Denver Bridge Infrastructure:
Rens, K. L., Nogueira, C.L., Neiman, Y., Gruber, T., and Johnson, L. (1999). Bridge management systems for the City and County of Denver. Pract. Period. Struct. Des. Constr., 4(4). 131-136.
Rens, K. L., Kim, T. Quebec Bridge Inspection Using Common Nondestructive and Destructive Testing. Journal of Performance of Constructed Facilities, 21(3).
Allen, B Nondestructive evaluation of 6th avenue viaduct using strain gauges. Denver, CO. Univ. of Colorado at Denver, 2002.
18


Allen, B., and Rens, K. L. (2004). Condition assessment of the east-bound 6th avenue viaduct using strain gauges. J. Perform Constr. Facil., 18(4), 205-212.
Gogel, M. E. Inspection and Rating System for Tubular Steel Pedestrian Bridges Denver, CO: Univ. of Colorado at Denver, 2001
Hager, A. Nondestructive evaluation of the westbound 6th avenue viaduct. Denver, CO: Univ. of Colorado at Denver, 2005.
Nogueria, C. Application of nondestructive evaluation techniques in bridge inspections as a tool for bridge management systems." Denver, CO: Univ. of Colorado at Denver, 1998.
Rens, K. L., Nogueira, C.L., and Transue, D. Bridge Management and Nondestructive Evaluation. Journal of Perfomance of Const. Facilities., 19(1). 3-16.
Rens, K. L., and Transue, D. (1999). Tomographic Imaging of Cracked Pier Cap of Evans over Santa Fe Bridge. Concrete Repair: International Concrete Repair Institute., 15(4). 22-29.
Rens, K. L., Transue, D., and Schuller, M. Acoustic Tomographic Imaging of Concrete Infrastructure: History, Interest and Applications. Journal of Infrastructure Systems, ASCE. 6(1). 15-23.
Wood. J and Rens, K. Nondestructive Testing of the Lawrence Street Bridge" ASCE Structures Congress, St Louis MO.
Rens, K., Nogueria C., and Transue, D. Bridge Evaluation Using Nondestructive Evaluation The Denver Management System. 3rd Forensic Congress, San Diego, California, ASCE press, 207-220.
1.8 CONCLUSION
The scope of work for each intern can be catered to the needs of the City and/or the needs of the student. The projects listed here are a sample of the variety of projects the University and the City have worked on together.
19


This thesis summarizes four evaluations completed as part of the bridge management project encompassed within the cooperative research program. Each structure presented distresses warranting an in-depth analysis to in order to determine the cause or best repair method. Chapter 2 discusses non-destructive evaluation techniques that are referred to in the evaluation summaries in the subsequent chapters. Chapter 3 summarizes the finite element analysis of the Quebec over Airlawn Bridge and repairs. Chapter 4 summarizes evaluation of cracking and leakage from the 20th Street HOV Bridge. Chapter 5 discusses the causes of cracking of the City of Cuernavaca Park Pedestrian Bridges and repair recommendations. Chapter 6 discusses the pier cap evaluation of the Evans over Santa Fe Bridge.
20


2. INVESTIGATION TECHNIQUES
2.1 INTRODUCTION
Structural evaluation requires detailed observations beyond that which is obtainable from design plans or as-builts. Field observations utilizing visual, destructive, and nondestructive techniques are required to obtain the information necessary to assess the extent of distresses and damages and to determine the remaining structural capacity of damaged elements. Often one needs to understand the condition of materials inside the structural member as well as the condition of the visually observable surfaces. Of the wide variety of non-destructive techniques available, a few were utilized for the purposes of this research. These techniques include visual inspection, sounding, ultrasonic velocity testing, and rebound hammer testing. Each one of these methods elicits valuable data, that when combined can paint an accurate picture of the condition of the structural material and provide insights into the best repair methods.
2.2 VISUAL INSPECTION
The first, and perhaps most important, non-destructive field inspection technique is visual inspection. Careful visual inspection can determine the course that the rest of the inspection may take and may lead one to re-assess preliminary assumptions. Different materials require different focuses for visual inspection. Concrete and steel were the primary structural elements observed for evaluation of the structures in this thesis.
21


Concrete is a material that has many mechanisms for deterioration: efflorescence is a deposit of salts on the surface of the concrete and is indicative of the passage of water through the concrete matrix. The presence of efflorescence indicates a potential for corrosion of the reinforcement and may qualify the porosity of the concrete. Cracks in concrete are indicative of tension forces that may signify underperformance or deterioration of the material. While hairline cracks are generally not structurally significant, wide cracks demand scrutiny as to their cause and extent. Delamination and spalling of concrete are often clear evidence of deterioration beyond the weathering surface of the material and may signify localized deterioration of encapsulated reinforcement or of the concrete itself. The presence of honeycombing or surface voids may indicate a lack of consolidation of concrete during placement, which can affect structural capacity.
Steel should be visually inspected for corrosion, which occurs as a result of oxidation of the iron within the steel. Corrosion is indicative of section loss which may affect the load carrying capacity of the member. Fatigue is failure of steel due to repeated stressing of the materials at levels below the yield limit of the member. Bridges are subject to classic fatigue loading. Deformation may be due to overloading, heat damage, or collision and should be examined for relevancy to structural performance.
In addition to visual inspection for materials defects, one must examine the interaction of the materials with each other and with applied loads. Camber, deflection, plumbness, orientation, and geometry should all be documented as part of the visual inspection.
22


2.3 SOUNDING
A variety of sonic percussion tools are employed to detect deterioration and delamination in concrete and timber elements. Hammer sounding is the process of tapping on the surface of the concrete with a hammer and listening for differences in the pitch of the sound made at impact. Generally, relatively lower pitches can be assumed to have lower stiffnesses which may be due to delamination of the concrete surface. These areas are often referred to as having a dull or hollow sound. Chain dragging, rotary percussion sounding devices, and steel rods are also used in the concrete sounding technique. ASTM D4580 is a standard developed for the assessment of bridge deck delamination using sounding techniques. Generally, delaminated sections are 'mapped' both on the concrete itself and on drawing representations of the concrete surface. The Schmidt hammer, to be discussed later, is a version of sounding as well.
2.4 ULTRASONIC VELOCITY
Ultrasonic velocity testing allows the observer to measure the quality of concrete throughout a solid member. It consists of applying two transducers to a concrete member at designated locations. One transducer produces an ultrasonic wave, the other receives the wave. Using the time difference between sending the signal and receiving it, in combination with the distance between the transducers, can be used to determine the velocity of the wave. High velocities are indicative of high quality, rigid concrete. Low
23


velocities indicate poor quality concrete. Table 2-1 illustrates a correlation of concrete quality to velocity.
Table 2-1 RELATIONSHIP BETWEEN CONCRETE QUALITY AND VELOCITY (Rens & Taewan,
2007)
Km/sec ln/(Js Quality of Concrete
>4.5 >0.1772 Excellent
3.5-4.5 0.1378-0.1772 Good
3.0-3.5 0.1181-0.1378 Doubtful
2.0-3.0 0.0787-0.1181 Poor
<2.0 <0.0787 Very Poor
Many different types of test equipment and testing methods have been examined with ultrasonic velocity. For the purposes of this report, P wave velocity was measured with 50 kHz transducers.
2.5 REBOUND HAMMER
The rebound hammer, also known as the Swiss hammer or Schmidt hammer, measures the surface hardness of concrete. The device propels a steel rod at the concrete with a calibrated amount of energy and measures the rebound energy. The rebound energy has been correlated with concrete compressive strength and is often used to determine the viability of loading of a fresh concrete pour. It is important to note that only the surface
24


hardness of concrete can be measured with this device. Results of rebound hammer testing
can be used in correlation with the hammer sounding technique to map surface concrete conditions.
2.6 TOMOGRAPHY
Ultrasonic velocity equipment can be utilized to create section cuts of a concrete member. The interior mapping of concrete deterioration is created by applying reconstruction algorithms that statistically evaluate multiple ray paths through a single section of concrete. This method has been proven effective for evaluation of concrete quality and for locating voids within a mass of concrete. As with ultrasonic velocity, continuous and uninterrupted concrete is required for accurate results.
2.7 CONCLUSION
Individually, the investigation techniques listed above elicit specific information about one aspect of a material. When many techniques are used and the data is taken in context with each other, one can begin to formulate an accurate description of a structural member's condition. This conditional data can be used to complete ratings on bridge structures as part of regular bridge infrastructure maintenance or can be used to help prioritize repairs and maximize maintenance economy for the managing agency.
25


3. QUEBEC OVER AIRLAWN BRIDGE
3.1 OVERVIEW
As part of regular bridge maintenance and inspection by the City and County of Denver (CCD), the Quebec over Airlawn Bridge was determined to be near structurally deficient as defined by the Federal Guidelines for Bridge Inspection. The University of Colorado Denver was asked to determine the feasibility of a deck repair in light of structural inconsistencies between the three adjacent bridges which make up the structure. Previous NDE studies and rehabilitation schemes on the substructure were proposed by Rens and Kim (2007).
3.2 LOCATION
The Quebec over Airlawn Bridge is located just south of the intersection of Quebec Street and Interstate Highway 70 over Airlawn Road. The spanned feature was originally a railroad spur but is now an out-of-service roadway. There are currently four lanes that pass over the roadway.
3.3 DESCRIPTION
The Quebec over Airlawn Bridge was built to span a now non-existent railroad spur and is a combination of three individual reinforced/prestressed concrete structures, of which the longitudinal axes are generally oriented in the north/south direction. The center structure
26


was originally built in 1961; the easternmost and westernmost structures were constructed
as part of widening efforts in 1971. The bridges were constructed at the behest of the State of Colorado (the State) under funding from the Federal Highway Administration. The bridges remained under the ownership of the State until 2005 when the State relinquished a portion of State Highway 35 to CCD ownership.
The center structure consists of three spans and is 58 feet (17.68 meters) wide; the north and south spans are approximately 48 feet (14.63 meters) long; the middle span is approximately 70 feet (21.33 meters) long. The structure was built on a 40 degree skew to accommodate the railroad spur's orientation relative to the road. The substructure consists of two reinforced concrete multiple column piers founded on spread footings and two reinforced concrete stub abutments founded on piles as is illustrated in Figure 3-1. The
Figure 3-1 VIEW OF SUBSTRUCTURE AND SUPERSTRUCTURE
27


superstructure consists of prestressed concrete AASHTO I-beams. The north and south span
girders cantilever approximately 5 feet (1.52 meters) into the center span. The center span is a suspended span connected by a hinge configuration to the cantilevered girders extending from the north and south spans. Diaphragms are located at abutments, piers, and midspans.
The easternmost and westernmost structures (side structures) were built directly adjacent to and along the same skew as the center structure. They are each 16 feet (4.88 meters) wide with the decks constructed structurally independent of the center bridge. Generally, with the exception of the superstructure, the side structures are consistent in orientation and design with the center bridge. The superstructure uses identical prestressed I-beams, but they are terminated over the piers rather than cantilevering over the piers as exists in the center structure. All three spans of the side structures are simple span configurations. Diaphragms are located at abutments, piers, and midspans.
As part of the construction of the side structures, a longitudinal joint was installed between the decks of the three bridges. The joint consisted of a polyvinyl carbonate tube that was cast in place with the deck and a pourable sealant was applied above it to complete the seal.
3.4 HISTORY
Upon acceptance and subsequent inspection of the structures in 2005, deterioration and delamination of the side structure's concrete pier caps became a concern. The caps
28


cantilever approximately 2-1/2 feet (76 centimeters) beyond the exterior column supports,
and due to the skew are partially exposed to direct precipitation and vehicle splash-back of water and deicing chemicals. Additionally, the transverse expansion joints above the pier had failed at the exterior edges allowing for direct drainage from the deck to top of the substructure element. Due to these factors, a study was completed to determine the structural condition of the caps.
Rens and Kim (2007)used destructive and nondestructive testing techniques to evaluate the condition of the pier caps. These techniques included: visual inspection, hammer sounding, rebound hammer, ultrasonic velocity, compressive strength, chloride content tests and petrographic analysis
As a result of the investigation, Rens and Kim determined that the cracking on the exterior faces of the caps were consistent with reinforcement cage locations, and as a result corrosion of the reinforcement had occurred. Generally, the deterioration of the concrete was limited to the exterior of the reinforcement cage. This conclusion was further supported by the destructive testing techniques. The deteriorated concrete indicated lower compressive strengths, higher chloride contents, and the presence of alkali-silica gel around the aggregate. The northwest pier cap was found to be in the most deteriorated condition; the southwest pier cap was in the best relative condition.
29


The likely repair, prior to testing, was to be complete removal and replacement of the pier
caps. However, as a result of the 2005 investigation, it was determined that the caps were structurally sound and that the repair should consist only of removal and replacement of the distressed exterior concrete cover.
3.5 PROBLEM
In addition to the repair of the distressed concrete, CCD examined the possibilities available for the repair of the deck joints. Not only had the transverse joints above side structure pier caps failed, but local failure of the longitudinal joints between the structures had been noted as well. Figure 3-2 shows spalling and efflorescence along the longitudinal joint. CCD
Figure 3-2 SPALLING AND EFFLORESCENCE ALONG LONGITUDINAL JOINT
30


suspected that the failure of the longitudinal joints was likely due to the internal concentrated stresses developed as a result of the elemental inconsistencies between the superstructures of the adjacent bridges.
The center structure had been designed with cantilevered north and south spans and a shorter suspended center span. The side structures had been designed as simple spans terminating over the pier centerlines. The differing span lengths, support, and boundary conditions, along with asymmetric loading due to lane placement, led to development of concentrated stresses in the connecting longitudinal joints, especially at the hinge locations. It was decided that replacement of the joint in kind would not be a viable repair option due to its short life expectancy.
Another option presented was to rigidly connect the reinforced concrete decks and eliminate the joint completely. In order to determine the viability of this repair, the stresses induced in rigid deck connection needed to be investigated. Due to the unique structural site conditions, it was determined that a Finite Element Analysis would be an appropriate method to achieve the desired results. An evaluation of existing differential movement between the structures would allow for an evaluation of the feasibility of this repair methodology.
31


3.6 ANALYSIS
Two analyses were completed in order to determine the stresses in a conjoined deck. The first was to examine the existing differential vertical deflections between the adjacent structures, the results of which are contained in this thesis. The second was to examine the
Figure 3-3 SUBSTRUCTURE PLAN VIEW
internal stresses induced within the conjoined deck itself, assuming a completed repair. The second analysis was performed by Jiang. Stress patterns within Jiang's model indicate high stress points at the hinge locations and above the piers in the center structure. Figure 3-3 illustrated the orientation of the adjacent structures in plan view.
Two models were constructed to determine the differential vertical deflections of the existing adjacent structures. According to the CCD, the shear connectors utilized in the design and construction of the bridges between the deck and prestressed girders were likely inadequate to transfer the design horizontal shear loads. As a result, a conservative model allowing for partial differential horizontal movement between the members was to be
32


Figure 3-5 COMBINED SUPERSTRUCTURE ELEMENT MODEL
Figure 3-4 SUPERSTRUCTURE SHELL ELEMENT MODEL
constructed. An additional model considering the composite action between the deck and prestress girders was also to be constructed and analyzed. The models were constructed using the finite element method software, SAP2000 v.10.0 (Computers and Structures, 2005). The first model was constructed using four node thin shell elements to model the
33


deck and two node frame elements to the model the superstructure and substructure as
shown in Figure 3-5. Shear connectors were provided between the deck and superstructure at spacing consistent with shear key spacing as detailed in the provided design documents. The stirrup shear connectors were neglected for the purposes of this model. The second model was constructed of four node thin shell elements representing the deck and superstructure and two-node elements representing the substructure as shown in Figure
3-4. Edge constraints were provided between the girders and deck thereby modeling a fully composite section. Bearing locations, hinge locations, and element and cross-section properties were consistent with data obtained from design documents and with data obtained in the field.
HS-20 truck and lane live loading were applied consistent with the AASFITO load resistance factor design prescriptive load application. Additionally, given the high truck volume over the structure and its proximity to traffic lights, a maximum loading simulating a continuous truck loading condition, was applied to the structures. Additional, considerations for construction loading sequences were also simulated to provide accurate results.
SAP2000 utilizes a moving load function in which live loads are incrementally applied to the structure to simulate movement across the bridge. The envelope reactions, deflections, and stresses are then provided for use in analysis and design. For the purposes of the analysis, critical points were identified for scrutiny. These points were chosen because they represented areas that would experience the highest levels of stresses under a joined deck
34


configuration due to differential superstructure deflection and rotation. The points adjacent
to the hinge locations and points at mid-span were chosen. Figure 3-1 identifies the critical points in plan view. For the purposes of this report, the south spans are span 1, the middle spans are span 2 and the north spans are span 3.

Max and Min Deformation Mid-spans
Rotation
- At and adjacent to center structure hinges
Figure 3-1 CRITICAL STRESS AREAS
35


3.7 RESULTS
Table 3-0-1 and
Table 3-0-2 show the deflections and rotations were calculated at the critical points.
Table 3-0-1 ANALYSIS RESULTS OF DIFFERENTIAL MOVEMENT OF FRAME AND SHELL
MODEL
Span 1 (Inches) (cm) Span 2 (Inches) (cm) Span 3 (Inches) (cm) CR1 (degrees) CR2 (degrees) CR3 (degrees) CR4 (degrees)
Max 0.1669 (0.434) 0.4923 (1.25) 0.3178 (0.807) 0.00259 0.00214 0.00154 0.00137
Table 3-0-2 ANALYSIS RESULTS DIFFERENTIAL MOVEMENTS OF SHELL MODEL
Span 1 (Inches) (cm) Span 2 (Inches) (cm) Span 3 (Inches) (cm) CR1 (degrees) CR2 (degrees) CR3 (degrees) CR4 (degrees)
Max 0.1296 (0.329) 0.5033 (1.28) 0.1832 (0.465) 0.00275 0.00323 0.00222 0.00212
The vertical deflection differentials were taken at the midspans and represented the envelope conditions of deflection for each structure. The numbers presented are essentially the difference between the highest upward deflection and the lowest downward deflection at the identified locations under the aforementioned applied loads. Likewise, the rotations
36


presented are the difference between the largest positive rotation and largest negative rotation.
Generally, both models resulted in calculation of similar envelope defections and rotations. The largest differential vertical deflections were observed in Span 2, at approximately 1/2 inch (1.27 cm). The largest differential rotations were observed at critical point 2, averaging approximately 0.003 degrees.
3.8 REPAIRS
Due to the extent of deterioration on the pier caps, rehabilitation was deemed necessary. CCD decided, based on the previous study by Rens and Kim (2007), that removal and replacement of the damaged concrete and reuse of the concrete core was a viable repair option. Consultants were employed to provide formal plans for repair. The scope of work includes replacement of the damaged concrete and installation of corbels under affected girders to distribute the bearing loads into the sound concrete.
Additionally, based on calculated deck stresses, a design was developed to tie the three concrete decks together monolithically. The east longitudinal joint has already been eliminated and the west longitudinal joint is slated to be monolithically tied at the onset of the pier cap repairs.
37


3.9 CONCLUSION
The Quebec over Airlawn Bridge presented unique challenges in analysis due to the structural differences in the adjacent superstructures. Long term deterioration of the longitudinal deck joints presented a threat to both the substructure and superstructure and required remediation. A finite element method model was created to measure the differential deflections and rotations currently being experienced between the superstructures.
Non-destructive testing and analytical analysis are often required to develop rehabilitation schemes. Fore example, a subsequent study of internal stresses within this structure allowed for the development of viable deck repair options.
38


4. 20th STREET HOV VIADUCT
4.1 OVERVIEW
During routine inspections for the State of Colorado, a wide crack was noticed in the north box of the 20th Street HOV Viaduct. The crack seemed to propagate along the encapsulated post-tension duct in the north web of the box. Additionally, large deposits efflorescence were observed to be leaching from the crack. The University of Colorado was asked to determine the origin of the leaching material and whether the prestress tendon had been adversely affected by the water infiltration. In other words, was the leaching a result of tendon deterioration?
4.2 LOCATION
The viaduct is located adjacent to and above 20th Street near downtown Denver as shown in Figure 4-2. For the purposes of this thesis, the structure is independent of the east ramps, originating west of Wynkoop Street and terminating at its intersection with Interstate 25. The focus of the investigation was the north side of the viaduct just to the east of where it overpasses Chestnut Street as shown in Figure 4-1.
39


Figure 4-2 LOCATION AND ORIENTATION OF 20TH STREET VIADUCT
Figure 4-1 VIEW OF SUPERSTRUCTURE AND SUBSTRUCTURE OF 20TH STREET HOV VIADUCT
40


4.3 DESCRIPTION
The viaduct was constructed in 1994 by the Regional Transportation District, the local public transit company, as part of the 20th Street Reconstruction Project. It is a 17 span cast-inplace, post-tensioned structure with a concrete composite deck. The intersecting ramps and one-lane sections of the structure are supported by a two box system, the two lane sections are supported by a three box system. The abutments and piers are comprised of reinforced concrete founded on a drilled pier foundation system. The spans average around 200 feet (60.06 meters) in length and are connected to each other by hinge assemblies. Shorter spans were constructed without the inclusion of prestressing strands. There were several vertical and horizontal curves designed and built into the structure.
The structure is maintained under the regulations of an intergovernmental agreement between the State of Colorado, the Regional Transportation District and the CCD. The agreement transferred complete ownership of the structure to the CCD. The CCD is to perform all major maintenance on the structure but the cost of the major maintenance is to be shared equally between the three parties. The snow removal and snow maintenance was to remain under the purview of RTD. RTD is also to perform all standard maintenance of the portion of bridge over the spanned DUT property. The State is to perform all standard and routine maintenance the portions of the structure not spanning the DUT property. As such, any deck removal or major rehabilitation of the structure, related to the findings here, would likely fall under the responsibility of the CCD.
41


4.4 HISTORY
The structure had been inspected several times prior to this investigation. The first inspection on record was completed in 1995. The following are excerpts from the element level summaries from the 1995 inspection (Lonco, 1995):
"104 4 P/S CONCRETE BOX GIRDER
HORIZ CRCKING IN WEBS ALONG P/S DRAPE AT VARIOUS LOCATIONS, SEE PHOTO. "
"23 5* BARE CONC W/EPOXY REBAR
EPOXY INJECTION REPAIRS WERE MADE AT SEVERAL LOCATIONS."
"358 4 DECK SURFACE CRACKING
RANDOM CRACKING THROUGHOUT, SEE PHOTO."
"359 5 CONC DECK BOTTOM/SOFFIT
LONGITUDINAL AND TRANSVERSE CRACKS W/EFFLORESCENCE, SEE PHOTO."
Note that this inspection was completed relatively shortly after construction; the cracking
along the prestressing strands was present even then. Also note the presence of
widespread longitudinal cracking in the deck.
The most recent inspection on record prior to this report was completed in 2004 (Inc., 2004). The following are excerpts from the element level inspection performed in 2004.
"104/1 P/S Cone Box Girder
CS3=Horizontal cracks .3mm to ,4mm in webs along P/S drape at various locations of girders 3A=12ft, 3B=25ft, 4B=63ft, 5B=156ft, 6A=43ft, 6B=79ft, 10C= 145ft, llA=18ft and HC=135ft, see photo. CS2=Horizontal cracks ,2mm in webs along P/S drape at various locations of girders 4A~126ft, 4B=10ft, 5A=46ft, 5b=l6ft, 7A-10ft, 7B=49ft, 8A=18ft, 8B=14ft and 9A=23ft,
42


10B=8ft, and 10C=2ft. 2ftx2ftx6inch void in interior web of 11C=CS2. Light cracking in end walls and diaphragms..."
"25/1 Cone Deck/Rig OL+Bars
Delaminated deck at various locations throughout the structure, approximately 5%, see photos. Spalls and voids at various locations, see photos. Moderate random cracking throughout, see photos."
"359/1 Soffit Smart Flag
Longitudinal and transverse cracks with efflorescence, see photos."
Although the extent of the horizontal cracks was not explicitly stated in the 1995 inspection, it can be assumed that the cracking was not considered significant due to the lack of detail included by the inspector and the rating of 'condition state 1' for the entire length of the prestress girders. In 2004, the inspection explicitly stated the extents and lengths of the horizontal cracking and the inspector dropped the rating for 1,006 feet (306.63 meters) of the prestress girder length; 682 feet (207.87 meters)were rated at 'condition state 3', 324 feet (98.76 meters) were rated as 'condition state 2'. According to the inspection reports, while the horizontal cracking was present shortly after construction, it became more extensive and severe over the course of 10 years. Similarly, the bare concrete deck rating was dropped from a 'condition state 1' in 1995 to a 'condition state 3' in 2004.
4.5 PROBLEM
Horizontal cracking seemingly following the drape of the prestress tendons, as illustrated in Figure 4-3, was a typical distress observed in the box girders throughout the structure. This was true of the north face of span 13 as well. The crack in the north web of this span,
43


however, was exhibiting significant deposits of effloresce along the line of the crack. The
CCD became concerned that the water transporting the efflorescence was originating from within the tendon duct, and thereby may be corroding the prestress tendon as well.
Figure 4-3 EFFLORESCENCE AT HORIZONTAL CRACK
Use of post-tensioning in concrete beams gained popularity in the 1950's and has seen widespread application since that time. The technology has been in use long enough to begin to see the long term effects and areas of concern through its service life. The State of Florida has done extensive research into the failure mechanisms of the technology (Transportation, 2003). A particular area of concern is insufficient grouting of tendon ducts and the subsequent collection of water. Standing water hastens the already accelerated deterioration of the tendons, known as stress corrosion. This fact lends validity to the concerns that the CCD had about the apparent leakage.
44


4.6 ANALYSIS
Visual inspection, rebound hammer testing, ultrasonic velocity testing, tomography, and chemical analysis were used to evaluate the condition of the box girder web and prestress duct. A representative region reflecting the most extensive deterioration was chosen for the evaluation.
4.6.1 SETUP
Access to the interior of the box girder was at the west end of the span near pier 12. Access ports between boxes were located at mid-span. A 6 inch (15.24 cm) by 6 inch (15.24 cm) grid was established on both sides of the web. The grid was 22 columns long and 5 rows high. A total number of 110 grid points were established for use in evaluation; the grid was 126 inches (320 cm) wide by 30 inches (76.2 cm) tall. The horizontal crack ran parallel to the horizontal lines of the grid and was approximately 8 inches (20.32 cm) from the bottom of the grid as shown in Figure 4-4.
45


Figure 4-4 GRID LAYOUT, BOLD LINE IS HORIZONTAL CRACK 4.6.2 VISUAL INSPECTION
From the interior of the box, a wide crack, approximately 37 feet (11.27 m) long, was observed in the bottom of the web near the mid-span. The orientation of the crack was consistent with the location and direction of the lower encapsulated tension steel duct. Efflorescence was observed to be accumulating along the bottom of the crack and even flowing down the web to the bottom flange in some locations. Additional efflorescence was observed along transverse cracks in the soffit/deck and traced down the web of the wall, terminating at the aforementioned horizontal crack. Voids created as part of construction were observed at the top flange-to-web interface. Areas around these voids were wet to the touch. There was no evidence of rust staining. Generally, the interior faces of the box were spotted with light honeycombing. Figure 4-5 shows a merged photographic
46


representation of the interior face at the distress region of the north box girder. The crack
has been highlighted in blue; the efflorescence has been highlighted in pink.
Figure 4-5 INTERIOR FACE OF THE NORTH WEB
From the exterior, a crack, approximately 18 feet (5.49 m) long, was observed to partially mirror the crack on the interior. Light efflorescence was observed near the center of the grid. Like the interior, the exterior face was pocketed with areas of honeycombing. Figure
4-6 is a merged photograph of the exterior face of the north web. The crack has been highlighted in blue; the efflorescence has been highlighted in pink.
47



*
A
*
%
Figure 4-6 EXTERIOR FACE OF THE NORTH WEB
4.6.3 REBOUND HAMMER
The rebound hammer uses measurements of rebound energy to determine the surface hardness of stiff material. Five measurements were taken at every grid location and then averaged to determine the concrete stiffness at each grid point. Unfortunately, the relevance of the rebound hammer testing is limited due to the prevalence of honeycombing on both the interior and exterior faces of the web. Figure 4-7 and Figure 4-8 show a topographic representation of the surface hardness of the concrete as a result of rebound hammer testing. The peaks represent stiff concrete; the valleys are relatively weak concrete. Figure 4-9 and Figure 4-10 show shaded versions of the surface hardness. The darker shade represents stiff concrete; the lighter shades are relatively weak concrete.
48


Figure 4-7 INTERIOR FACE REBOUND HAMMER RESULTS TOPOGRAPHIC
Figure 4-8 EXTERIOR FACE REBOUND HAMMER RESULTS TOPOGRAPHIC
49


Figure 4-9 INTERIOR FACE REBOUND HAMMER RESULTS SHADED
Figure 4-10 EXTERIOR FACE REBOUND HAMMER RESULTS SHADED 4.6.4 ULTRASONIC VELOCITY TESTING
Ultrasonic velocity testing measurements were taken at each location. Direct through shots were taken from one grid point to the corresponding grid point on the opposite face of the web. Additionally, a shot was taken from each grid point to the adjacent grid points (up to 3 points, or 18 inches[45.72cm], away) on the opposing face.
50


KvXv
Hr
>c
li
Figure 4-11 SIDE VIEW OF ULTRASONIC VELOCITY RAY PATHS
Figure 4-11 shows a side view section cut of the analysis area. The crossing lines display the path of the ultrasonic velocity measurements. Figure 4-12 shows a top view section cut of the analyzed section. Again, the crossing lines display the path of the ultrasonic velocity measurements.

Figure 4-12 PLAN VIEW OF ULTRASONIC VELOCITY RAY PATHS
Using the direct method (shot directly from one grid point to the corresponding grid point on the opposite face only) the quality of the concrete was determined. Figure 4-13 shows
51


the quality of the concrete graphically. The darkest shade is poor quality concrete (velocity
of 0.0787 in/ps [0.20 cm/ps]- 0.1181 in/ps [0.30 cm/ps]). The lightest shade is excellent concrete (velocity of >0.1772 in/ps [0.45 cm/ps]). Generally, the concrete is of poorer quality near the bottom of the grid and of better quality near the sides and top of the grid.
Figure 4-13 DIRECT METHOD CONCRETE QUALITY
Tomography yielded similar results. Figure 4-14 shows a tomographic section cut through the lower prestress duct. In this image, dark blue represents good concrete and red represents poor concrete. The tomography generally shows medium to good concrete throughout the entire grid. No voids were detected within the prestress duct and all poor concrete seemed to be on the outer perimeter of the duct. This may be due to poor consolidation as is evidenced by the honeycombing on the exposed faces of the web.
52


Figure 4-14 TOMOGRAPHIC SECTION CUT THROUGH LOWER PRESTRESS DUCT 4.6.5 CHEMICAL ANALYSIS OF EFFLORESCENCE
As part of the investigation, samples of the efflorescence were chemically analyzed for content. The samples were retrieved via the scraping method from the inside face of the distressed web area. The samples taken were white to translucent, dry, powdery, and with geometric grains. The samples were tested for chloride, iron, and zinc content.
The samples were tested for chlorides to determine if the leakage contained chemicals that would hasten the corrosion of mild or stressed reinforcement. The test for iron was to
53


provide evidence of existing corrosion, should any be found present. The zinc analysis would indicate the passage of the water over and through the galvanization of the prestess ducts.
The possible contributors to each of these chemicals were identified as drainage, contained mild reinforcement or stressed steel, ingredients of the concrete, and deicer chemicals.
EXECUTIVE SUMMARY - Detection Highlights
D7C160372 REPORTING ANALYTICAL
PARAMETER RESULT LIMIT UNITS METHOD
N19E1 03/12/07 12:00 001
Chloride 150 B,G 600 mg/kg SW846 9056
N19K2 03/12/07 12:00 002
Iron 320 L 15 mg/kg SW846 6010B
Zinc 6.7 3.0 mg/kg SW846 6010B
S24W1 03/12/07 12:00 003
Chloride 270 B,G 600 mg/kg SW846 9056
S24W2 03/12/07 12:00 004
Iron 330 L 15 mg/kg SW846 6010B
Zinc 14 3.0 mg/kg SW846 6010B
Figure 4-15 CHEMICAL ANALYSIS RESULTS 20TH STREET HOV VIADUCT
The chemical analysis results are presented in Figure 4-15. The laboratory reported that a chemical was present within the sample matrices that inhibited the detection of chlorides.
54


This interference within the samples was not a common error according to the testing agency. Among the reported possible interfering chemicals were chemicals used as rust inhibitors.
Chemical analysis revealed that the iron levels were much higher than would be expected even for heavy corrosion combined with high iron content cement and fly ash. The zinc levels were consistent with that of tap water.
The chemical levels were compared to all the aforementioned possible contributors. They most closely matched the chemical content of the deicing chemicals used by the Colorado Department of Transportation. The deicing chemicals contain very high levels of iron and magnesium as well as the expected high levels of chlorides. Additionally, a rust inhibitor additive is included to prevent the corrosion of cars and utility lines around the roadways. It is decided that the inhibitor was the matrix interference in the chloride tests.
4.7 CONCLUSION
Longitudinal cracks corresponding to encapsulated prestress ducts are a common distress among post-tensioned structures. Concrete box girders are especially vulnerable because they are often used in high torsion situations. The cracks are caused by a combination of stresses developed around the duct due to the prestressing forces. It is often difficult to properly consolidate the concrete around multiple ducts, which leads to less cross-sectional
55


area to withstand the required stresses. According to the Federal Highway Administration's
Post Tensioning Installation and Grouting Manual,
"The distance between the outside of the duct and the side of the web should be adequate to accommodate the vertical reinforcing and specified cover and provide the minimum concrete section to satisfy design requirements."
While the design for this structure did specify a 1-1/2 inch (3.81 cm) gap between the ducts, evidence of honeycombing and improper consolidation may have resulted in the observed cracking.
No significant voids were detected within the duct by the ultrasonic velocity tomography and there was no evidence of rust staining within the efflorescence. Furthermore, the chemical analysis of the efflorescence revealed a chemical composition similar to what would be expected to be originating from above the deck. Based on the aforementioned analysis and observations, it is concluded that the origin of the observed seepage and efflorescence is external chemicals due to deicing applications and subsequent bleeding through the deck cracks and accumulating along the transverse crack.
56


5. CITY OF CUERNAVACA PARK PEDESTRIAN BRIDGES
5.1 OVERVIEW
Despite previous repairs, the City of Cuernavaca Park pedestrian bridges manifested distress in the form of cracking on the backwalls of the abutments. Previous repairs included insertion of a concrete expansion joint, patching and providing for ample expansion and contraction movement of the railing system. The University was asked to determine the cause of the cracking so a repair could be formulated to permanently alleviate the distresses.
5.2 LOCATION
The City of Cuernavaca Park is located to the east of Interstate Highway 25 between 20th Street and Park Avenue near lower downtown Denver as shown in Figure 5-1. The park is
Figure 5-1 OVERVIEW OF CITY OF CUERNAVACA PARK
57


bisected by the South Platte River which runs in a general south to north direction. Two
pedestrian bridges were constructed to span the river to accommodate foot traffic through and within the park.
5.3 DESCRIPTION
The two pedestrian bridges were oriented in a curvilinear pattern to accommodate an elliptically shaped pedestrian path within the park as shown in Figure 5-2. The bridges were
*= *-
Figure 5-2 ELLIPTICAL LAYOUT OF BRIDGES AND FOOT PATH
comprised of three simple spans, horizontally curved on an approximate 500 foot radius. Each span varied in length but all were within the range of 98 feet (29.87 m) to 118 feet (35.97 m) long. Its superstructure is comprised of two parallel tube steel subdivided warren trusses with a reinforced concrete deck. The substructure is comprised of two reinforced
58


concrete abutments and two reinforced concrete piers and all are founded on H-pile
foundations. Figure 5-3 show the abutment from the side.
Figure 5-3 PHOTO OF PEDESTRIAN BRIDGE ABUTMENT
The design intended each span to have a pinned-roller support condition. A sole plate was welded to the bottom chord of the trusses and imbedded in the bearing seats. No direct means of rotation was provided for the pinned assemblies. The roller assemblies consist of a steel sole plate welded to the bottom chord and made to bear directly onto a steel masonry plate that is embedded into the concrete. The sole plate was provided slots and a bolt was installed within the slots to laterally restrain the superstructure.
59


5.4 HISTORY
The bridges were designed in 1996 by the Sheflin Group, a structural engineering firm in Denver. The park and structures were constructed in 1996 as part of the rehabilitation of the lower downtown area.
In 2000, inspectors noticed that the bridge abutments were manifesting cracks at the intersection of the backwall and the wingwall. According to the inspection report (Lonco I., 2002), "West abutment cracked from rail thermal expansion." The CCD became concerned when the cracking developed into local spalling and lateral displacement of the backwall. Upon the recommendations of a consultant, repairs were completed to mitigate the distresses. The concrete and reinforcement between the backwall and wingwall was severed and a bond-breaker was installed between the abutment elements. The cracks and spalls were cosmetically repaired to match the original condition. Finally, the approach rail attachment brackets were slotted to allow for thermal movement of the rails.
5.5 PROBLEM
In 2003, inspectors noticed that the cracking and spalling of the abutments had again manifested. The distresses seemed to exactly replicate the distresses observed prior to the previous repairs. UCD was asked to re-evaluate the structure and determine the cause in order to formulate repairs to permanently mitigate the distresses.
60


5.6 VISUAL INSPECTION
Visual inspections of the deck, superstructure and substructures were completed to determine the in-situ condition of the structures. The conditions reported below are considered typical for both structures.
The bearings were examined for condition and evidence of functionality. As described above, the moveable bearings consisted of a steel plate to steel plate assembly. Visual examination determined that corrosion between the plates had occurred. There was no evidence of sliding or wear that would be present should the bearing be allowing movement as designed. The bearing guide bolts were also examined for movement. At two locations it was discovered that the guide bolts had only been imbedded approximately % inch (1.27 cm) into the concrete and could be easily removed by hand. All the bolts were, however, vertical indicating that only negligible pressure had ever been applied to them prior to the inspection. Figure 5-4 shows an expansion bearing assembly. Figure 5-5 shows a typical fixed bearing assembly and Figure 5-6 shows a non-functioning guide assembly.
61


Figure 5-4 MOVEABLE BEARING ASSEMBLY
Figure 5-5 TYPICAL FIXED BEARING ASSEMBLY
62


Figure 5-6 MISSING GUIDE BOLT IN GUIDE ASSEMBLY
Figure 5-7 TYPICAL VERTICAL CRACK IN ABUTMENT BACKWALL
63


Vertical cracks were observed on the exposed face of the backwall of the abutment as
shown in Figure 5-7. The cracks were consistent with the cracks pictured in the inspection reports prior to the previous repairs. The cracks were located at the backwall to wingwall intersections.
Spalling, as shown in Figure 5-8, was observed adjacent to the expansion joints at the abutment. The edges of the spalls seemed to saw-cut, indicating that the previous repair materials had failed to properly bond. Figure 5-9 displays the crack orientation on the backwalls of the abutment.
Figure 5-8 TYPICAL CRACKING AT ABUTMENT CORNER
64


Figure 5-9 LOCATION OF CRACKS ON BACK WALL
5.7 ANALYSIS
Many potential problems with the structures were examined for their effects on the abutments. First, load transfer between the deck and abutment was considered. Second, thermal expansion of the concrete and geometric restraints within the abutment itself was
65


examined as a possible source of internal stresses. Lastly, interaction of the superstructure
and the substructure via the bearing devices was considered.
In order to accurately model the stress distribution and interaction, the bridges were constructed within SAP2000 v.10, a finite element method software. Linear elastic analysis was completed to model stress distributions in all models.
Several load cases were considered for use in the analysis of the structures and diagnosis of the deterioration. As always, live loading, dead loading, snow loading, and thermal loading combinations were first considered. Sustained loading such as dead loading and cyclic thermal loading became the primary suspects due to the slow reoccurrence of the cracking despite interim repairs. For the purposes of this report, both the live and snow loading were viewed as contributing factors rather than as direct causes of the deterioration.
5.7.1 MODEL 1-BRIDGE AND DECK TRUSS
The bridge deck and truss system was modeled according to geometry as described in design documents provided by the City. The 118 foot (36 m) span of the south bridge was chosen because it is the longest span, it interacts with the abutment and it is a good representation of typical superstructure geometry. The concrete deck was modeled using solid elements and the steel truss system was modeled with frame elements. Figure 5-10 shows the model in three dimensions.
66


Upon completion of the model, a number of analyses were run to determine the possible
reactions the abutment would supply to the superstructure under different support conditions.
First, total seizure of the roller bearings above the piers was modeled by pinning both ends of the span at the bearing points.
Second, a support condition assuming partial seizing of the slide bearings above the piers was modeled by allowing movement of the bridge with rollers and springs to partially limit the movement.
Third, the bridge was modeled using rollers that freed tangential movement and restricted all other translation above the piers.
Last, complete horizontal translational movement was modeled such that the direction and magnitude of unrestricted thermal movement above the pier could be ascertained.
In all cases, the end of the span that interacts with the abutment was pinned as per the original design. When longitudinal movement was permitted, it was modeled as tangential to the radius of curvature. This tangential modeling was consistent with the orientation of the bearing slots as observed on the structure.
67


Figure 5-10 SUPERSTRUCTURE AND DECK MODEL
The superstructure model was subjected to two long-term loading conditions in order to obtain the reactions. First, it was subjected to the dead loads due to the self weight of the structural elements of the bridge. Second, as per AASTO specifications, it was subjected to thermal loading through a temperature range of -40F to 120F. The loads were modeled concurrently for the purposes of this study.
68


5.7.2 MODEL 2: CONCRETE BRIDGE ABUTMENT
Model 2 is a model of a typical abutment for the structures and was constructed as described in the design documents. The model in its entirety consists of solid, eight point elements.
As part of the previous repair, the holes in the base plates of the rail supports were reamed out so as to allow for movement of the rail independent of the abutment. For this reason, the rails were neglected as they could not have induced or transferred stresses into the abutment.
In order to model the structure's soil interaction, springs were applied at all nodes that interface with the soil. The springs were given spring constants equivalent to passive soil pressure. A typical granular fill was assumed because specific geotechnical information was not available for review at the time of this analysis.
This model was used a template from which all other abutment models were constructed. Loadings, both dead and thermal were applied to this model. The assumed temperature range was between -40F and 90F per AASFITO specifications.
69


Figure 5-11 BRIDGE ABUTMENT MODEL
70


5.8 Model 3 H-Pile
As previously mentioned, the abutment lies on grade and was attached to H-piles near its four corners. The H-piles were embedded into bedrock approximately 30 ft below grade.
Initial models of the abutments disallowed movement and rotation of the concrete where it attaches to the H-piles to simulate the pile's supportive properties. Upon further consideration, it was decided that the stiffness of the H-piles may play a role in the structural element interactions of the abutment under thermal loading of the superstructure.
Several assumptions were made in order to best simulate the interaction between the H-piles and the soil, thereby allowing their stiffness to affect the structure. When the H-piles do not push into the soil, they are loaded by active soil pressures which provide lateral support for the piles. When the piles push laterally into the soil, the soil reacts by pushing back with passive soil pressure
The first assumption was that the piles sit vertically such that any loading from above was predominately axial. In reality, this was not always the case. As H-piles were driven into the ground, they frequently bend and veer according to the geologic composition and strata.
The exact path of any given pile was not known for the purposes of this study.
71


The second assumption is that the piles were driven through a single homogeneous soil from beginning to end. There was no record of geotechnical exploration related to this structure and, therefore, a continuous soil condition per regional soil records was assumed. According to the United States Geological Survey, the soil in this area is comprised of elolian deposits defined as sand and silt. For the purposes of this study mixed sand and silt characteristics were used to define the soil properties.
The third assumption was the interaction between the piles and the abutments was monolithically joined. Any structural degradation that may be present sub-grade had not been noticed or examined by any inspectors and therefore would not be considered in this analysis.
Finally, elastic springs were used to laterally support the piles. They were assigned constants consistent with equivalent fluid passive soil pressures. The ends of the piles were fixed for a distance of 6 feet to simulate the bedrock embedment. Figure 5-12 shows the model including the H-pile supports.
72


Figure 5-12 ABUTMENT MODEL INCLUDING STIFFNESS OF H-PILE FOUNDATION
73


5.9 RESULTS
The presence of an expansion joint at an abutment is usually indicative of a moveable bearing condition below. In these structures, however, the bearings at the abutments were all fixed bearings. The fact that the deck was connected to the abutment via an expansion joint was likely an architectural rather than structural detail.
Stresses within the abutment model were calculated under thermal loading. Stresses were observed to be concentrated at the locations where cracking had occurred. Tension stresses as high as 600 psi (4137 kPa) were calculated in the face of the backwall. Figure
5-13 and Figure 5-14 show the stress distributions under positive and negative thermal loading, respectively.
74


Compression Tension
Figure 5-13 ABUTMENT STRESSES UNDER POSITIVE THERMAL LOADING
75


Compression Tension
Figure 5-14 ABUTMENT STRESSES UNDER NEGATIVE THERMAL LOADING
76


As previously described, the superstructure model was subjected to thermal loading under different support conditions. The pinned connections modeled above the piers were evaluated for horizontal reaction such that the reaction forces could be applied to the abutment model at the appropriate locations. In order to properly model the interaction, the bearing locations were allowed to laterally expand per the calculated lateral expansion of the abutment seat. Table 5-1 indicates the largest reactions imposed into the abutment based on the different bearing assumptions.
Table 5-1 ENVELOPE REACTIONS BASED ON BEARING ASSUMPTIONS
Outer Radius Bearing Inner Radius Bearing
Fixity Rx Ry Rz Rx Ry Rz
Pin-Pin -275 k 56.4 k 50.7 k -174 k 4.23 k 20.5 k
Pin-Springs -73 k 49.9 k 52.6 k 21 k -31 k 18.5 k
Pin-Rollers 52.2 k 43.9 k 53.7 k -52 k -37 k 17.4 k
The maximum reactions were then applied to the abutment model to measure the stresses induced into the abutment. While significant stresses were measured, the distribution did not match the crack patterns observed in the abutment. However, these stresses were additive when superimposed with the thermal stresses at the crack locations.
Figure 5-15 and Figure 5-16 show the stress distribution under positive and negative thermal loading, respectively, in combination with the pin-pin reactions, show the stress distribution under positive and negative thermal loading, respectively, in combination with the pin-pin reactions
77


Compression Tension
Figure 5-15 POSITVE THERMAL LOADING SUPERIMPOSED WITH PIN-PIN REACTIONS
78


Compression Tension
Figure 5-16 NEGATIVE THERMAL LOADING SUPERIMPOSED WITH PIN-PIN REACTIONS
79


Compression Tension
Figure 5-17 POSITIVE THERMAL LOADING SUPERIMPOSED WITH PIN-SPRING REACTIONS
80


Compression Tension
Figure 5-18 NEGATIVE THERMAL LOADING SUPERIMPOSED WITH PIN-SPRING REACTIONS
81


Compression Tension
Figure 5-19 POSITIVE THERMAL LOADING SUPERIMPOSED WITH PIN-ROLLER REACTION
82


Compression Tension
Figure 5-20 NEGATIVE THERMAL LOADING SUPERIMPOSED WITH PIN-ROLLER REACTIONS
83


Figure 5-17 and Figure 5-18 show the stress distribution under positive and negative thermal loading, respectively, in combination with the pin-spring reactions. Figure 5-19 and Figure 5-20 show the stress distribution under positive and negative thermal loading, respectively, in combination with the pin-roller reactions.
5.10 CONCLUSION
In summary, the thermal loadings produced concentrated tension stresses likely large enough to cause slight tension cracking. However, the stresses induced via the bearings into the abutment by the thermal loading of the superstructure in combination with the thermal loading of the substructure far exceed the tension capacities of the concrete. It is likely that this loading interaction caused the cracking in the backwall of the abutment.
The previous repairs addressed this cause by eliminating the rigid joint between the backwall and wingwalls. However, the joint was patched over with a rigid concrete material thereby eliminating the ability for the joint translate.
According to the FEM analysis, the superstructure, given the ability freely translate in the horizontal plane; when one end is fixed, the other end moves in a linear path away from the fixed location. The bearings were placed such that movement would be allowed only tangentially to the curvature of the bridge. Even if the bearings were functioning as designed, the stresses induced into the abutment under thermal loading would likely still rupture a rigid connection in the backwall of the abutment.
84


5.11 REPAIRS
The cracking at the expansion joint that was rigidly patched should have been expected. New patches should terminate at the joint such that stresses induced during thermal expansion and contraction are not transferred into the repair. Additionally, new elastomeric bearings should placed under the superstructure to allow for thermal expansion and contraction as was the original intent of the design.
85


6. EVANS OVER SANTA FE BRIDGE
6.1 OVERVIEW
Despite previous repairs, the Evans over Santa Fe Bridge manifested distresses in form of cracking and severe delamination on bent cap beam number 6. The University was asked to determine the cause of the damage so a repair could be formulated to permanently alleviate the distresses.
6.2 DESCRIPTION
The Evans over Santa Fe Bridge was constructed in 1971. The structure carries Evans Avenue and spans Delaware Street, railroad lines and Santa Fe Drive. It is approximately 785 feet long with six 80 foot (24.38 m) spans and three 95 foot (28.96 m)spans. There are two approaches and four ramps connecting to the structure. The structure was designed with unique geometry to accommodate a single-point intersection near the center of the bridge. Figure 6-1 shows a plan view of the structure.
86


Figure 6-1 PLAN VIEW OF EVANS OVER SANTA FE BRIDGE
The superstructure was comprised of steel I-beams and a compositely connected concrete deck. The substructure consists of reinforced concrete bents and abutments. The superstructure bears on the bents via a elastomeric bearing pads. The bents consist of a bent cap beam supported by 5 round concrete columns.
There are several plank-seal type expansion joints integrated into the design of the superstructure. The expansion joints were located above the piers and abutments to accommodate the simple span design.
87


6.3 HISTORY
In 1995 inspections to the structure indicated severe deterioration to the concrete below the expansion joints above bent #6, bent #3 and both abutments. The deterioration had manifested in the form of severe delamination, spalling, cracking, and reinforcement corrosion. The structure was deemed repairable and the city developed repair plans to address the issues.
In general the repair consisted of removal of the damaged concrete by mechanical means.
A bonding agent was then applied to the newly exposed reinforcement and concrete surfaces. The concrete member was reformed and patched to match the original extents of the concrete faces. A final coat of Sikagard, a concrete sealant and protective coating, was applied all the concrete surfaces. Figure 6-2 is an excerpt from the 1995 repair documents.
88


re* W**T CAP l*UM.% mvttm muxii nu
m swim
TYPICAL BENT CAP REPAIRS SECTION
dostno comi amo ncpronciMo
TYPICAL COLUMN REPAIRS
SLCHflN
Figure 6-2 PLANS FOR REPAIR OF DELAMINATED AREAS
In 2002 a wide vertical crack was observed extending through the bent cap at the cantilevered south end of bent #3. Rens and Transue (2002)completed non-destructive testing on the end cap of the bent beam to determine the extent of propagation of the
89


crack. Testing revealed that the crack did not extend through entire cross-sectional area. The end cap was closely monitored by the city following the investigation.
In the summer of 2008 the city was notified that a large section of concrete had fallen from the bent cap. Subsequent inspection revealed that a portion of the bearing area of one of the girders had spalled away. The remaining concrete was observed to be delaminated and cracking. A temporary repair was immediately installed to re-support the girder and to reopen all traffic lanes.
6.4 PROBLEM
The extent of distress to the pier cap was obscured by the sealant coating applied during the previous repairs. An investigation into the depth of delamination and extent of cracking was required to determine structural adequacy and to formulate repairs.
6.5 ANALYSIS AND RESULTS
Visual inspection, sounding, rebound hammer testing, and ultrasonic velocity testing including tomography were used to evaluate the condition of the cap beam. A 4-inch (10.16cm) by 4 inch (10.16cm) grid was established on the beam to demarcate testing locations for purposes of analysis. The grid extended around the entire perimeter of the cap beam. At the centerline of the bridge the grid was 48 inches (1.2 m) high. Near the end caps the grid extended 44 inches (1.12 m) vertically.
90


6.5.1 VISUAL INSPECTION
A large crack was observed extending across the south end cap of the beam. The crack had visually widened as compared to photographs from the previous non-destructive testing completed in 2002. Figures 6-3 and 6-4 compare the cracks in the south pier cap.
Figure 6-3 EAST FACE OF SOUTH PIER CAP
Figure 6-4 WEST FACE OF SOUTH PIER CAP
91


Evidence of water staining was observed above the exterior columns on the east face of the beam. Water stain was observed above the south column on the west face of the beam. The water staining is indicative a failure of the overlying expansion joint.
The south end of both the east and the west faces was 'caked' in a white translucent deposit. Based on experience, the deposits resembled the residue left by multiple applications and subsequent evaporation of deicing chemicals.
Additional cracking plate cracking consistent with severe delamination was observed throughout the surface of the cap beam. Associated peeling of the Sikagard sealant observed in areas near the aforementioned water staining.
6.5.2 SOUNDING
The cap beam was sounded with a 3 pound (6.6 kg) hammer around the perimeter of the member. Figure 6-3 and Figure 6-4 illustrate the delaminated areas of the east face and west face, respectively. Overall, the southwest region of the cap beam exhibited the most delamination. The center of the member was in the best relative condition.
r W ' * y, r*

6-5 DELAMINATION ON EAST FACE OF BENT #6
92


6-6 DELAMINATION ON WEST FACE OF STRUCTURE
6.5.3 REBOUND HAMMER
The rebound hammer uses measurements of rebound energy to determine the surface hardness of stiff material. Five measurements were taken at every grid location and then averaged to determine the concrete stiffness at each grid point. The rebound hammer testing yielded measurements highly corollary with the sounding method.
6.5.4 ULTRASONIC VELOCITY INCLUDING TOMOGRAPHY
For purposes of comparison, both end caps were evaluated using ultrasonic velocity testing. The south end cap could not be fully evaluated due to ineffectual wave propagation past the aforementioned crack. As discussed in Chapter 2, a continuous, non-interrupted concrete member is required to measure wave velocity through concrete. It became apparent that the crack had fully transgressed the member at that location. Figure 6-7 and Figure 6-8 display tomographic representations of the end caps where measureable.
93


4-inch
Figure 6-7 TOMOGRAPHIC REPRESENTATION OF THE NORTH END CAP
4-inch
24-inch
44-inr.h
Figure 6-8 TOMOGRAPHIC REPRESENTATION OF THE SOUTH END CAP
94


Attempts were made to create tomographic section cuts near the delaminated sections of concrete. However, the severity of the delamination disallowed the propogation of ultrasonic waves through the medium making measurement of the velocity impossible.
6.6 CONCLUSION
In summary, the severe delamination was observed throughout many areas of the cap beam. Delaminated sections disallowed sonic wave propagation limiting the effectiveness of the ultrasonic velocity measurement. Tomographic section cuts were produced at each column head and end caps.
The delamination observed on the cap beam is generally consistent with the delamination observed prior to the 1995 repairs. The depth of the delamination is consistent with recommendations made in the repair notes, generally to the depth of the reinforcement cage. Figure 6-9 and Figure 6-10 show comparisons of the repairs versus the current condition of the caps.
95


Figure 6-9 PREVIOUS REPAIR COMPARED TO EXISTING DELAMINATION OF WEST FACE
Figure 6-10 PREVIOUS REPAIR COMPARED TO EXISTING DELAMINATION OF EAST FACE
96


When comparing the end cap tomography with that obtained by Rens and Transue in analysis completed in 2002, it is apparent that the crack has continued to propagate. In 2002, the signal path was sent and received on opposing sides of the crack. The condition upon this investigation was such that no signal could penetrate the substrate past the crack. Because of this fact, exact extent of propagation is difficult to ascertain. The delamination on the sides of the cap near the end cap added to the difficulties in obtaining through shots. Based on the inability to shoot through the member it is prudent to assume that the end cap has separated from the cap beam such that the encapsulated steel alone transfers the stresses from the overlying girder. The tributary area affecting this girder line is significant when considering that it supports the cantilevered section of the deck and partially stabilizes the overlying parapet in a vehicle collision.
97


7. SUMMARY AND CONCLUSIONS
The evaluation of four bridge structures located in the City and County of Denver were summarized in this report. The evaluations were completed as part of a cooperation between the CCD bridge infrastructure management division and the University of Colorado Denver. The program has assisted the municipality with inspection and evaluation of a large portion of the City's infrastructure and given relevant experience to a number of students and faculty.
The Quebec over Airlawn Bridge was exhibiting deterioration as a result of leakage from a failure of an overlying longitudinal joint. The structure was modeled via a finite element program to determine the design load flexure the current geometry allows for. This information was used to determine the feasibility of monolithically connecting the decks of the adjacent structures. The repairs, including the monolithic joining of the decks, are slated to occur in 2009.
The 20th Street HOV Viaduct was evaluated to determine the origin and effects of leakage observed from within the box girders. Non-destructive testing was completed to determine whether voids were present in the post-tension ducts, the contents of which are sensitive to corrosion. Chemical analysis was performed on the deposits of efflorescence to identify its potential origin. No unwanted voids were detected in the ducts within the region evaluated. Furthermore, chemical analysis indicated that the leakage had originated from
98


deicing chemicals through cracks within the deck, not from the deterioration of the tendon duct as suspected.
The City of Cuernavaca Park Pedestrian Bridges were observed to be manifesting distress at the backwalls of the abutments. Despite addressing similar distress with previous repairs, the cracking continued to propagate. Finite element modeling of the structure yielded significant stresses from thermal loading and observations indicated that the expansion bearings were not properly aligned to allow for longitudinal expansion and contraction. It was concluded that the combination of these factors led to the observed distress. The University recommended installation of new bearings that would allow the structure to move as designed.
The Evans over Santa Fe Bridge was exhibiting severe delamination and cracking on a bent cap beam under an expansion joint. Non-destructive testing was completed to determine the extent of damage to the beam. Sounding and tomography yielded a high correlation between previous repairs and the current extent of damages. It was concluded that the repairs previously performed on the beam had failed to properly adhere leading to the existing concrete condition which subsequently led to further deterioration. Furthermore, upon comparison with previous NDE studied, it was determine that the crack on the south end cap of the cap beam had propagated completely through the member.
99


A. APPENDIX A CRACK MAPPING OF 8th AVENUE VIADUCT PIERS A.l PIER 2
3810*
r-----------------------------------------------------------------1
NORTH SOUTH
SOUTH NORTH
Figure A-l PIER 2 CRACK MAP
100