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Non-destructive evaluation of the westbound Sixth Avenue viaduct

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Title:
Non-destructive evaluation of the westbound Sixth Avenue viaduct
Creator:
Hager, Angela S
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English
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xiv, 293 leaves : ; 28 cm

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Subjects / Keywords:
Nondestructive testing -- Colorado -- Denver ( lcsh )
Viaducts -- Testing -- Colorado -- Denver ( lcsh )
Nondestructive testing ( fast )
Viaducts -- Testing ( fast )
Sixth Avenue (Denver, Colo.) ( lcsh )
Colorado -- Denver ( fast )
Colorado -- Denver ( fast )
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bibliography ( marcgt )
theses ( marcgt )
non-fiction ( marcgt )

Notes

Bibliography:
Includes bibliographical references (leaves 291-293).
General Note:
Department of Civil Engineering
Statement of Responsibility:
by Angela S. Hager.

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|University of Colorado Denver
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Auraria Library
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All applicable rights reserved by the source institution and holding location.
Resource Identifier:
62873299 ( OCLC )
ocm62873299
Classification:
LD1193.E53 2005m H33 ( lcc )

Full Text
NON-DESTRUCTIVE EVALUATION OF THE
WESTBOUND SIXTH AVENUE VIADUCT
by
Angela S. Hager
B.S., Colorado School of Mines, 2003
A thesis submitted to the
University of Colorado at Denver
in partial fulfillment
of the requirements for the degree of
Master of Science
Civil Engineering
2005


This thesis for the Master of Science
degree by
Angela S. Hager
has been approved
by
Chengyu Li
g/z/bf
Date


Hager, Angela Susan (M.S., Civil Engineering)
Non-Destructive Evaluation of the Westbound Sixth Avenue Viaduct
Thesis directed by Associate Professor Kevin L. Rens
ABSTRACT
This thesis presents the procedure and results of a non-destructive series of
testing completed on the westbound 6th Avenue Viaduct in Denver, Colorado. The 6th
Avenue Viaduct was modified in a 1997-1998 rehabilitation project making the entire
superstructure continuous. New sliding bearing pads, installed at this time, failed to
properly accommodate expansion and contraction of the bridge superstructure due to
temperature changes, thus generating excessive loadings in the pier and foundation
elements. Over time, this contributed to the deterioration of the bridge. Excessive
rotation about the bridge longitudinal axis has spalled the concrete in the pier
columns, raising concerns as to the longevity of the bridge. The bridge deck does not
have sufficient capacity to meet the desired inventory rating. Various forms of testing
were required to determine the extent of damage, the depth and necessity of repair, to
revise ratings, and to determine the optimal procedure for a proposed repair. Thus,
the deck, three of the reinforced concrete pier columns and the composite piles of the
111


west abutment necessitated non-destructive insitu testing to determine condition and
strength.
The testing provided valuable information for the repair of the structure. The
testing of the deck reassured the City and County of Denver that they would not have
to replace the entire deck. Rather, testing identified a localized area with a
substandard rating to be repaired. Based on the testing of the pier columns, all of the
piers will need to be replaced. The two piers within the railroad clearances will need
to be built utilizing a crash wall, or of heavy construction. Otherwise the City must
obtain variance from the railroad. The testing of the fixed pier indicated that its
replacement should be scheduled to occur during the first stage in the repair program.
Lastly, the testing of the composite piles, coupled with visual observations, suggested
that replacement was not necessary provided that the longitudinal joint be replaced.
The results of testing presented within this thesis, provided the City and County of
Denver with valuable information for determining the repair needs of a structure that
is in severe need of rehabilitation.
This abstract accurately represents the content of the candidate's thesis. I recommend
its publication.
Signed
IV


ACKNOWLEDGMENT
I would like to acknowledge and thank the following people who assisted in the work
required for the study.
University of Colorado at Denver
Kevin Rens
Sam Brown
Mike Doyle
Paul Bountry
City and County of Denver
Jim Barwick
Jim Hamblin
William Melton
Bill Cusack
URS Corporation
Chengyu Li
Craig Parent
Special thanks to my teacher, friend and mentor, Kevin Rens for his continual
guidance and support. Additionally, particular thanks to Jim Barwick and the City and
County of Denver for funding this research.


CONTENTS
List of Figures......................................................... x
List of Tables.......................................................... xiv
Chapter
1. Introduction....................................................... 1
1.1 Cooperative Research Background.................................... 5
1.2 Sixth Avenue Viaduct Background.................................... 6
1.2.1 1997-1998 Sixth Avenue Rehabilitation.............................. 9
1.2.2 The Eastbound Bridge Studies and Modifications.................... 11
1.2.3 The Westbound Sixth Avenue Viaduct................................ 13
1.3 Scope of the Project.............................................. 21
1.4 Testing Procedure................................................. 23
2. Previous Research................................................. 25
2.1 Bridge Testing.................................................... 25
2.2 Non-Destructive Testing Methodologies............................. 28
2.2.1 Visual Inspection................................................. 32
2.2.2 Hammer Sounding................................................... 33
2.2.3 Rebound Test Hammer............................................... 34
2.2.4 Ultrasonic Pulse Velocity......................................... 36
2.2.5 Tomography........................................................ 40
2.2.6 Concrete Core Specimen Removal.................................... 41
2.2.7 Core Compression Testing.......................................... 43
2.2.8 Core Modulus of Elasticity Testing................................ 45
VI


2.2.8.1 Strain Gauge Installation Procedures.............................. 51
2.2.9 Core Shear Testing................................................ 55
2.2.10 Rebar Tensile Testing............................................. 58
2.2.11 Steel Thickness Testing........................................... 60
3. Testing of the Deck............................................... 63
3.1 Deck Steel Reinforcement Tensile Testing.......................... 64
3.2 Deck Concrete Core Compression Testing............................ 67
3.3 Deck Concrete Core Modulus of Elasticity Testing.................. 69
3.4 Deck Core Shear Testing of White Topping Layer.................... 70
3.5 Deck Concrete Core Ultrasonic Pulse Velocity Testing.............. 72
3.6 Deck Testing Conclusions.......................................... 73
4. Testing of the Pier Columns....................................... 76
4.1 Pier Columns Visual Inspection.................................... 79
4.2 Pier Columns Hammer Sounding...................................... 80
4.3 Pier Columns Rebound Hammer Testing............................... 84
4.4 Pier Columns Direct Transmission Ultrasonic Pulse Velocity Testing. 88
4.5 Pier Columns Horizontal Tomography................................ 94
4.6 Pier Columns Vertical Tomography................................. 106
4.7 Pier Columns Combined NDE Testing................................ 118
4.8 Pier Columns Concrete Core Removal............................... 122
4.9 Pier Columns Concrete Core Compression Testing................... 124
4.10 Pier Columns Concrete Core Modulus of Elasticity Testing......... 126
4.11 Pier Columns Concrete Core Ultrasonic Pulse Velocity Testing... 127
4.12 Pier Columns Testing Conclusions................................. 128
5. Testing of the West Abutment Composite Piles..................... 132
vii


5.1 West Abutment Composite Piles Steel Thickness Ultrasonic Pulse
V elocity T esting............................................ 133
5.2 West Abutment Composite Piles Conclusions....................... 135
6. Possible Sources of Errors...................................... 136
6.1 Visual Inspection Sources of Error.............................. 137
6.2 Hammer Sounding Sources of Error................................ 138
6.3 Rebound Test Hammer Sources of Error............................ 139
6.4 UPV Direct Transmission Sources of Error........................ 141
6.5 Surfer Software Sources of Error................................ 143
6.6 Tomography Sources of Error..................................... 144
6.7 Core Specimen Removal Sources of Error.......................... 145
6.8 Core Compression Testing Sources of Error....................... 146
6.9 Core Modulus of Elasticity Testing Sources of Error............. 147
6.10 Core UP V Testing Sources of Error.............................. 149
6.11 Shear Testing Sources of Error.................................. 150
6.12 Rebar Tensile Testing Sources of Error.......................... 151
6.13 Composite Pile Steel Thickness Testing Sources of Error......... 152
6.14 Error Conclusions............................................... 153
7. Summary of Testing Results and Repair Recommendations........... 154
7.1 Deck Testing Conclusions........................................ 154
7.2 Pier Columns Testing Conclusions................................ 156
7.3 West Abutment Piles Testing Conclusions......................... 159
7.4 Proposed Repairs................................................ 160
viii


Appendix
A. Deck Modulus of Elasticity Data and Stress-Strain Diagrams....... 163
B. Pier Column Hammer Sounding Testing Data......................... 166
C. Pier Column Schmidt Hammer Testing Data.......................... 170
D. Pier Column UPV Direct Transmission Testing Data................. 174
E. Pier Column Horizontal Tomography Testing Data................... 178
F. Pier Column Vertical Tomography Testing Data..................... 209
G. Pier Column Modulus of Elasticity Stress-Strain Diagrams......... 234
H. CCD Deck Testing Report.......................................... 239
I. CCD Pier Column Core Sample Testing Report....................... 246
J. CCD Pier Column NDE Testing Report............................... 250
K. CCD West Abutment Composite Pile Testing Report.................. 255
L. Quebec over Sand Creek Bridge Testing Report..................... 260
Bibliography........................................................... 291
IX


LIST OF FIGURES
Figures
1.1 Sixth Avenue Viaduct Substructure Westbound (Left) Eastbound
(Right)............................................................ 7
1.2 Westbound Sixth Avenue Viaduct..................................... 8
1.3 Eastbound Sixth Avenue Viaduct..................................... 8
1.4 Typical Retrofit Hinge with Plates Welded to Flanges and Under the
Hanger Bar........................................................ 10
1.5 Flexural Cracking of Pier 17 of the Eastbound Sixth Avenue Viaduct.. 11
1.6 Typical Concrete Spalling of Westbound 6th Avenue Viaduct Pier
Columns........................................................... 18
1.7 West Abutment of the Westbound 6th Avenue Viaduct................. 20
2.1 Rebound Test Hammer............................................... 35
2.2 Ultrasonic Pulse Velocity Testing................................. 37
2.3 Ultrasonic Pulse Velocity Transmission Modes (a) direct; (b) semi-
direct; (c) indirect.............................................. 38
2.4 Concrete Core Drilling Equipment.................................. 41
2.5 Compression Testing Machine....................................... 44
2.6 The P-3500 Digital Strain Indicator (left) and SB-10 Switch and
Balance Unit (right).............................................. 46
2.7 Typical Strain Gauges............................................. 48
2.8 Wheatstone Bridge Schematic....................................... 49
2.9 Materials for Gauge Preparation and Installation.................. 51
2.10 Double Shear Test Block........................................... 56
X


2.11 Double Shear Test.................................................. 56
2.12 Force Diagram of Double Shear Test................................. 57
2.13 Epoch III Ultrasonic Flaw Detector................................. 61
2.14 Steel Ultrasonic Testing Display................................... 61
3.1 Location of Rebar Sample #1 (Exposed & Corroded)................... 64
3.2 Rebar Sample #1.................................................... 64
3.3 Location of Rebar Samples #2 and #3 (Near Pier 17 North Edge). 65
3.4 Core Drilling of Westbound 6th Avenue Deck......................... 67
3.5 Modulus of Elasticity Testing of Deck Core D6...................... 69
3.6 Deck Core Layers................................................... 70
3.7 Deck Core Shear Test Break......................................... 71
4.1 Identification Grid System......................................... 78
4.2 Identification Grid Layout......................................... 78
4.3 Pier 17 Crack Encapsulating Data Point El8......................... 79
4.4 Pier 11 Hammer Sounding Testing Results............................ 81
4.5 Pier 13 Hammer Sounding Testing Results............................ 82
4.6 Pier 17 Hammer Sounding Testing Results............................ 83
4.7 Pier 11 Schmidt Hammer Testing Results............................. 85
4.8 Pier 13 Schmidt Hammer Testing Results............................. 86
4.9 Pier 17 Schmidt Hammer Testing Results............................. 87
4.10 Pier 11 UPV Direct Transmission Testing Results.................... 89
4.11 Pier 13 UPV Direct Transmission Testing Results.................... 90
4.12 Pier 17 UPV Direct Transmission Testing Results.................... 91
4.13 Pier 17 UPV Direct Transmission Testing around the E18 Data Point.. 93
4.14 Pier 11 Horizontal Tomography Slice at Level 0................... 95
4.15 Pier 11 Horizontal Tomography Slice at Level 2................... 95
4.16 Pier 11 Horizontal Tomography Slice at Level 4................... 96
XI


4.17 Pier 11 Horizontal Tomography Slice at Level 6..................... 96
4.18 Pier 11 Horizontal Tomography Slice at Level 8..................... 96
4.19 Pier 11 Horizontal Tomography Slice at Level 10.................... 97
4.20 Pier 11 Horizontal Tomography Slice at Level 12.................... 97
4.21 Pier 11 Horizontal Tomography Slice at Level 14.................... 97
4.22 Pier 11 Horizontal Tomography Slice at Level 16.................... 98
4.23 Pier 11 Horizontal Tomography Slice at Level 18.................... 98
4.24 Pier 13 Horizontal Tomography Slice at Level 0..................... 98
4.25 Pier 13 Horizontal Tomography Slice at Level 2..................... 99
4.26 Pier 13 Horizontal Tomography Slice at Level 4..................... 99
4.27 Pier 13 Horizontal Tomography Slice at Level 6..................... 99
4.28 Pier 13 Horizontal Tomography Slice at Level 8.................... 100
4.29 Pier 13 Horizontal Tomography Slice at Level 10................... 100
4.30 Pier 13 Horizontal Tomography Slice at Level 12................... 100
4.31 Pier 13 Horizontal Tomography Slice at Level 14................... 101
4.32 Pier 13 Horizontal Tomography Slice at Level 16................... 101
4.33 Pier 13 Horizontal Tomography Slice at Level 18................... 101
4.34 Pier 17 Horizontal Tomography Slice at Level 0.................... 102
4.35 Pier 17 Horizontal Tomography Slice at Level 2.................... 102
4.36 Pier 17 Horizontal Tomography Slice at Level 4.................... 102
4.37 Pier 17 Horizontal Tomography Slice at Level 6.................... 103
4.38 Pier 17 Horizontal Tomography Slice at Level 8.................... 103
4.39 Pier 17 Horizontal Tomography Slice at Level 10................... 103
4.40 Pier 17 Horizontal Tomography Slice at Level 12................... 104
4.41 Pier 17 Horizontal Tomography Slice at Level 14................... 104
4.42 Pier 17 Horizontal Tomography Slice at Level 16................... 104
xii


4.43 Pier 17 Horizontal Tomography Slice at Level 18 Neglecting
Readings at E18.......................................... 105
4.44 Pier 17 Horizontal Tomography Slice at Level 18 Including
Readings at El8.......................................... 105
4.45 Pier 11 Vertical Tomography Line AE....................... 106
4.46 Pier 11 Vertical Tomography Line BF....................... 107
4.47 Pier 11 Vertical Tomography Line CG....................... 108
4.48 Pier 11 Vertical Tomography Line DH....................... 109
4.49 Pier 13 Vertical Tomography Line AE.................... 110
4.50 Pier 13 Vertical Tomography Line BF....................... Ill
4.51 Pier 13 Vertical Tomography Line CG.................... 112
4.52 Pier 13 Vertical Tomography Line DH....................... 113
4.53 Pier 17 Vertical Tomography Line AE.................... 114
4.54 Pier 17 Vertical Tomography Line BF....................... 115
4.55 Pier 17 Vertical Tomography Line CG....................... 116
4.56 Pier 17 Vertical Tomography Line DH....................... 117
4.57 Pier 11 Overall NDE Testing Results......................... 119
4.58 Pier 13 Overall NDE Testing Results......................... 120
4.59 Pier 17 Overall NDE Testing Results......................... 121
4.60 Top of Pier Column 17 East Side........................ 123
5.1 Steel Thickness Testing of West Abutment Composite Piles. 134
xiii


LIST OF TABLES
Tables
2.1 Correlation of Pulse Velocity with Concrete Quality................. 38
3.1 Deck Lower Level Transverse Steel Reinforcement Bars Tensile Test
Results.......................................................... 65
3.2 Original Concrete Deck Compression Testing Results.................. 68
3.3 Double Shear Testing Results........................................ 71
3.4 Deck Core Ultrasonic Pulse Velocity Testing Results................. 72
4.1 Pier Column Compressive Strength Testing Results.................. 124
4.2 Pier Column Modulus of Elasticity Testing Results (x 106 psi)... 126
4.3 Pier Column Concrete Core UPV Testing Results..................... 127
XIV


1. Introduction
Approximately 30% of the 600,000 bridges in the U.S. highway system are classified
as being either functionally obsolete or structurally deficient. [Hadavi, 1998; Tang,
2003] This classification does not mean that the deficient bridges are necessarily
unsafe; they are classified administratively to indicate that they require some form of
maintenance or major rehabilitation to restore them to their original condition or to
their current load carrying capacity. With the average age of the bridges in the US
Interstate system being 45 years, significant portions of the nations bridges are
presently approaching a condition state that will necessitate significant repairs,
rehabilitation, or even replacement [Tang, 2003].
The City and County of Denver (CCD) owns and maintains approximately 750
bridges located throughout an approximate area of 155 square miles. Of these bridges,
those with a low sufficiency rating and those with visual signs of deterioration may
prompt more in-depth studies to accurately assess their condition state as well as to
decide between repair alternatives. Of the CCDs major structures (those greater than
20 ft. in length) approximately 15 have a sufficiency rating less than 60, 4 bridges are
structurally deficient and approximately 50 are functionally obsolete. The Westbound
Sixth Avenue Viaduct, with a sufficiency rating of 54.6 and a functionally obsolete
rating is among the worst structures in the City. The purpose of this thesis study is to
utilize knowledge learned through several previous nondestructive evaluation (NDE)
bridge projects to accurately assess one of these bridges, the Westbound 6th Avenue
viaduct. Additionally, similar testing was completed to assess the condition of the
Quebec over Sand Creek bridge, and the analysis of this bridge can be found in
Appendix L.
1


This thesis presents the theory, procedure, results and conclusions of a series on non-
destructive tests completed on the Westbound Sixth Avenue Viaduct during the
summer of 2004. The testing completed for this study was utilized in the
development of a detailed scope of work for a possible repair project for the bridge.
Specifically, the results were used to determine repair necessity, decide between
repair alternatives and to establish an optimal construction sequence for specific
components of the bridge. By utilizing non-destructive evaluation techniques, the
bridge remained open and functional while providing the specific condition data
needed to evaluate its condition.
The Westbound Sixth Avenue Viaduct was originally constructed in 1958 and has
since undergone numerous repairs and rehabilitations. The viaduct was significantly
modified in a 1997-1998 rehabilitation project making the entire superstructure
continuous. New sliding bearing pads, installed at this time, failed to properly
accommodate expansion and contraction of the bridge superstructure due to
temperature changes, thus generating excessive loadings in the pier and foundation
elements. Over time, this contributed to the damage and deterioration of the bridge.
Additionally, excessive rotation about the bridge longitudinal axis has spalled the
concrete in the pier columns, raising concerns as to the longevity of the bridge.
Furthermore, the bridge deck rating, based upon as-built specifications, does not meet
the desired inventory rating. Therefore, various forms of testing were required to
determine the extent of damage, the depth and necessity of repair, to revise ratings,
and to determine the optimal procedure for the proposed repair of the bridge. Thus,
the deck, three of the reinforced concrete pier columns and the composite piles of the
west abutment were identified as necessitating non-destructive insitu testing to
determine condition and strength.
2


Testing of the deck consisted of concrete core compression, core modulus of
elasticity, core ultrasonic pulse velocity, shear testing of the white topping layer, and
rebar tensile testing. The results of the testing were utilized to modify the deck rating.
This modified deck rating isolated a location along the southern edge of the bridge as
being the only location with a substandard inventory rating.
A consulting engineer had recommended the replacement of all of the pier columns.
However, two of these piers are within the specified railroad clearance requirements.
These necessitated further testing of insitu conditions to determine strength
parameters to verify the necessity of replacement, and to decide between repair
alternatives. Also, the fixed pier for the bridge would need to be evaluated for loading
conditions throughout the repair. Testing of the pier columns consisted of a visual
inspection, hammer sounding, rebound test hammer, ultrasonic pulse velocity direct
transmission, and vertical and horizontal tomography. Additionally, core samples
were removed and tested for compressive strength, modulus of elasticity, and
ultrasonic pulse velocity. This testing raised specific concerns as to the structural
integrity of each pier. Therefore, the proposed repair project includes the replacement
of all of the piers, starting with the fixed pier. The two piers located within the
railroad clearance requirements will need to be uniquely designed to meet the railroad
construction requirements or variance from the railroad will need to be sought.
Finally, the piles of the west abutment appeared severely corroded raising concerns
that section loss may have compromised their strength. Therefore, steel ultrasonic
testing was completed to determine the amount of section loss in the steel of the
composite piles. These results indicated that only the south pile had undergone some
section loss but that it was not overstressed. Testing results coupled with visual
observations confirmed that replacing the longitudinal joint between the bridges
would prevent further damage to this pile.
3


The totality of the testing provided valuable information for the repair of the
structure. The testing of the deck reassured the City and County of Denver that they
would not have to replace the deck; rather, a viable means of increasing the deck
rating along the southern edge of the bridge was needed. Based on the testing of the
pier columns, all of the piers will need to be replaced. The two piers within the
railroad clearances will need to be built utilizing a crash wall, or of heavy
construction, otherwise the City must obtain variance from the railroad company. The
testing of the fixed pier indicated that its replacement should be scheduled to occur
during the first stage in the repair program. Lastly, the testing of the composite piles,
coupled with visual observations, suggested that replacement was not necessary
provided that the longitudinal joint be replaced. The testing and results presented
within this thesis provided the City and County of Denver with valuable information
for determining the repair needs of a structure that is in severe need of rehabilitation.
4


1.1 Cooperative Research Background
The University of Colorado at Denver (UCD) has been involved in cooperative
research efforts with the City and County of Denver (CCD) since 1997. Through this
relationship, UCD has been active in the inspection, rating, maintenance, and repair
of CCD infrastructure such as bridges, streets, alleys, curbs, and sidewalks. [Usagoni
2004, Santhantip 2002, Gogel 2001, Allen 1999, Nogueira 1998, Liephart 1994,
Rens et al, 1999a, Rens et al 1999b, Rens et al, 2003 Allen et al, 2004; Rens et al
2004; Rens et al 2002; Rens et al 1999;] The CCD Department of Public Works
owns, inspects, and maintains approximately 640 bridges in its inventory. The health
and condition of each of these bridges is fundamental to the overall transportation
system for the City of Denver. As a result of a variety of previous projects UCD has
completed on CCD bridges, UCD has developed Non-Destructive Evaluation (NDE)
experience. The purpose of this project is to utilize bridge-testing methodologies
established through several previous NDE bridge projects to provide an assessment of
various components of the westbound 6th Avenue Viaduct. The combined knowledge
obtained through this project will be provided to the CCD to accurately assess the
overall condition of the westbound 6th Avenue Viaduct. This will determine
recommendations for repair and rehabilitation of the bridge.
5


1.2 Sixth Avenue Viaduct Background
The 6th Avenue Viaduct consists of three separate structures. The two main bridges
(eastbound and westbound) each carry traffic in one direction over Union Pacific
Railroad (UPRR) and Regional Transportation District (RTD) Light Rail property as
shown in Figure 1.1. The third bridge, located just west of the main westbound
bridge, is a single span bridge. This bridge connects the viaduct to US Highway 6, an
indispensable highway in the Denver metro region. The viaduct facilitates
transportation between the western suburbs and Interstate 25 to the major areas of
commerce and employment throughout central Denver. The east side of the viaduct
connects to 6th Avenue, a major arterial street for City of Denver.
The viaduct consists of two westbound bridges and one eastbound bridge. The main
westbound bridge, as shown in Figure 1.2, is a 19-span structure built in 1958. The
substructure of this main westbound bridge consists of 18 single, circular reinforced
concrete piers with steel pier caps. The superstructure consists of six wide-flange
steel girders with a composite concrete deck. The smaller westbound bridge is a
single span structure that carries the 6th Avenue westbound traffic over Osage St and
was built in the mid-1960s. Finally, the eastbound bridge, as shown in Figure 1.3,
was built between 1963 and 1965, with an additional traffic lane and girder line added
in 1973. The eastbound bridge has a predominantly steel substructure with eight wide
flange girders and a composite concrete deck. For more information on the eastbound
structure, refer to Allen (2002) and Allen and Rens (2004). The main westbound
bridge, hereafter referred to simply as the westbound bridge, is the focus of this study.
6


Figure 1.1: Sixth Avenue Viaduct Substructure Westbound (Left) Eastbound (Right)
7


Figure 1.2: Westbound Sixth Avenue Viaduct
Figure 1.3: Eastbound Sixth Avenue Viaduct
8


1.2.1 1997-1998 Sixth Avenue Rehabilitation
Both the eastbound and westbound bridges originally utilized a system of simple,
overhanging, and suspended spans for the superstructure. Girders spanned between
two piers with overhangs on each end. Between adjacent overhangs, pin and hanger
bars held a suspended span in place. The hinge at each pin transferred shear forces,
but no moment. An expansion joint in the bridge deck, at each hinge joint,
accommodated differential movement between separate sections of the bridge, as well
as thermal movement of the superstructure.
Unfortunately, this original structural configuration of the bridges did not adequately
accommodate increased loads resulting from general increased use and from larger,
heavier vehicles. Most notably, the cantilevered and suspended portions of the bridge
girders allowed for an unacceptable level of live load deflection at several points on
the bridge. The level of deflection created a bouncy feel that was noticeable and
disconcerting to motorists. As bridge engineer Jim Barwick stated, when the viaduct
was designed, engineers used the standard joint of the time, which basically
consisted of a couple pieces of steel laid over each other and as anyone who has
driven over them knows, the hunks of metal stick out of the asphalt like railroad
tracks. [Wilmsen, 1997] Concerns about the serviceability, general deterioration of
the superstructure, and safety concerns regarding the pin and hanger bar connections
prompted the City of Denver to initiate a rehabilitation of both bridges that was
completed in 1998.
The primary purpose of this 1998 modification was to improve the service condition
of the bridge by eliminating expansion joints, and to increase the superstructure load
rating by making the longitudinal beams continuous. The 38 expansion joints were
9


replaced with just four, one at each end of each bridge. In order to achieve
continuity, the deck was cut at each of the original expansion joints and the beam
hinges were fixed by the addition of steel flange and web cover plates, as shown in
Figure 1.4. New expansion bearings were installed at the abutments and on all pier
columns except for the fixed piers of each bridge (pier 11 of the westbound bridge
and pier 8 of the eastbound bridge). The anchor bolts connecting the transverse steel
plate girders to the columns were cut at all piers except the fixed piers in order to
install the expansion bearings.
Figure 1.4: Typical Retrofit Hinge with Plates Welded to
Flanges and Under the Hanger Bar
The superstructure of the eastbound bridge is very similar to that of the westbound
bridge, while the substructures are quite different. These differences in the
substructures precipitated a much different response in the structural behavior by each
bridge. While signs of distress revealed some of the structural problems, analytical
work was required to determine the short-term safety and long-term viability of each
of the bridges.
10


1.2.2 The Eastbound Bridge Studies and Modifications
The eastbound 6th Avenue Viaduct is 1,430 feet in length and is supported by a series
of 17 piers. The bridge piers are numbered consecutively from east to west. Piers 1
through 16 are steel piers, built up from wide flange sections, consisting of either
straight or A-frame piers. Both the straight and A-frame steel piers have a pier cap
consisting of a wide flange section welded at the top. Unlike the other piers of the
eastbound bridge, Pier 17 is constructed of concrete with a rectangular concrete pier
cap. Eight wide flange girders of varying size bear on the top of each pier cap.
As the 1997-1998 rehabilitation project approached completion, cracking was
observed on Pier 17, the concrete pier of the eastbound bridge, as shown in Figure
1.5.
Figure 1.5: Flexural Cracking of Pier 17 of the
Eastbound Sixth Avenue Viaduct
Inspectors then performed an inspection of the bridge and observed that the new
sliding bearing pads installed in the 1998 rehabilitation project had not moved from
their installation positions. It was concluded that the bearing pads failed to properly
11


accommodate expansion and contraction of the bridge superstructure due to
temperature changes. Visible damage due to contraction of the superstructure in the
winter of 1999-2000 prompted a study of substructure elements to determine the
short-term safety and long-term viability of the bridge.
A non-destructive evaluation of the eastbound bridge consisted of a strain gauge
study to determine the overall health of the eastbound 6th Avenue Viaduct bridge.
Concrete and steel elements were instrumented with a total of 62 strain gauges and 2
concrete crack gauges, and data was taken for a period of three months. Results of
the study suggested that the bridge was safe for short-term use, but that the long-term
viability of the structure had been compromised. Repair or replacement of
substructure elements was recommended, along with frequent monitoring during the
interim period. [Allen, 2002; Allen and Rens 2004]
Due to the significant cost, the repairs were to be phased over time, dealing with the
most severe safety issues first. Repairs completed to date include:
Total replacement of Piers 11, 13, 15, 16, and 17
Foundation augmentation at Piers 8-17
Replacement of all bearing devices
These repairs have significantly improved the condition and lengthened the life
expectancy of the eastbound 6th Avenue bridge. As the eastbound repairs neared
completion, the City and County of Denver began to look to repair needs of the
westbound bridge.
12


1.2.3 The Westbound Sixth Avenue Viaduct
The Westbound 6th Avenue Viaduct carries 3-lanes of traffic between Osage Street to
Kalamath Street, providing a major link from downtown Denver to Interstate 25 and
further west. This 19-span bridge with a total length of 1341.5 feet was built in the
late 1950s and modified in the 1997-98-rehabilitation project. The original
superstructure consisted of 6 parallel, longitudinal rolled steel beams that were bolted
to transverse steel plate girders at each pier location. Several hinges in the beams and
expansion joints in the deck were originally placed at various locations along the
bridge to allow for temperature expansion. A series of 18 reinforced concrete
columns, numbered consecutively from east to west, support the transverse steel plate
girders. Anchor bolts embedded into the concrete attached the transverse steel plate
girders to the column. A single 4-foot diameter reinforced concrete column is used at
piers 1 through 10 and a single 4.5-foot diameter reinforced concrete column is used
at piers 11 through 18. A footing and four 2-foot diameter caissons support the pier
columns. A concrete spread footing supports the east abutment wall and the west
abutment consists of a concrete cap supported by steel pipe piles filled with concrete.
In addition to changing the continuity of the bridges, several other changes were
made to both bridges during the 1998 rehabilitation project. The additional changes to
the westbound bridge included changes to the curb and the deck surface. The curb on
the southern edge of the deck was removed and the curb on the northern edge was
replaced with Colorado Department of Transportation (CDOT) Bridge Rail Type 4.
The deck was milled and a concrete topping was cast on the existing deck concrete. A
2-inch asphalt layer was then paved on top of the concrete topping.
13


The expansion bearings installed in the 1998 modification, consisted of two stainless
steel plates welded to the sole plates, two recessed polytetrafluroetheylene (PTFE)
sheets vulcanized to the base plate, a base plate, and a plain elastomeric pad on top of
the concrete column. The base plate and the bearing pad were anchored to the
concrete column by eight anchor bolts. Uplift restrainer bars were attached to each
end of the sole plate. The stainless steel plates and the PTFE sheets form two sliding
surfaces one near the southern edge of the bearing and one near the northern edge of
the bearing. The distance between the centers of the two sliding surfaces is only 4
feet.
The bearings develop a tension-compression force couple to resist unbalanced
transverse rotational movements from heavy trucks traveling along the outer edges of
the deck. Because the distance between the tensile component and the compressive
component of this couple is only 4 feet, relatively large forces are developed in
response to the transverse moment. The large compressive force is transferred to the
top of the column, thus resulting in cracking and spalling of the concrete under the
bearing. In recent years, inspectors have increasingly observed this severe concrete
cracking and spalling on the concrete columns at these locations.
While such signs of distress revealed some of the structural problems, analytical work
was required to determine the load rating of the bridge. Therefore, in 2002, a
consulting engineering firm performed analytical work to determine the load rating of
the bridge. In so doing, they determined, among other things that the concrete deck
was rated 21.6 tons at inventory level, falling below the desired inventory rating of 36
tons.
14


These results coupled with visual observations of the condition prompted a field
inspection and monitoring program performed by the CCD, a consulting engineering
firm, and UCD, which provided the following findings:
The superstructure had moved to the north transversely about one inch, thus
resulting in contact between the bearing plates and the guide bars on the two
abutments. This movement allowed the longitudinal joint between the
eastbound and westbound bridges to open, thereby allowing water and other
elements to pass through the joint material. In addition, the movement induced
transverse shear forces along the abutment expansion joints for which the
joints were not designed to handle.
Excessive deformation had occurred in the elastomeric pads, indicating an
overstress condition. Contact was observed between the sole plates and base
plates on the expansion bearings. This contact caused by rotation of the
superstructure and deformation of the steel sole plates, induced dry friction
between the two steel surfaces, and increased the horizontal force transferred
to the pier columns and foundations.
Concrete at the top of the pier columns was vertically cracked on all of the
piers. Concrete sounding indicated that delamination might have occurred in
the cracked areas.
In addition to the aforementioned vertical cracks, severe concrete spalling was
observed on the northern faces of the tops of the columns at Piers 2, 4, 6, and
7. A significant area of concrete supporting the bearings had been lost due to
the spalling. The cause of the cracking and spalling may be attributed to the
increase of pressure on the north edge due to the transverse rotation of the
superstructure. Given the severity of the situation, temporary shoring towers
were erected at Piers 2, 4, 6, and 7 in January of 2001.
15


Additional vertical cracks in the pier columns were observed during the
construction of the eastbound bridge rehabilitation. As a means to prevent
further spalling of the concrete, a 5-foot length of steel casing was installed at
the tops of the columns during the summer of 2004. The casing was installed
at Piers 1, 3, 9, 12, 13, 17, and 18.
Concrete spalling was present at the ground level on Piers 1, 3, and 7. Severe
corrosion was observed on the exposed reinforcing steel. It is likely that
corrosion has occurred in the other piers.
In October of 2000, soil cracking and settlement had been observed around
the foundation of Pier 1. A monitoring program was implemented to measure
the column and footing movement at this pier. The field data obtained by
UCD during February to May of 2001 indicated that both the column and the
footing were rotating in the longitudinal direction of the bridge as temperature
fluctuated. The analysis of the data indicated that the column was overloaded.
The concrete cap at the west abutment displayed severe deterioration,
including concrete spalling and reinforcement corrosion. The elastomeric pads
showed extensive deformation.
Concrete sounding indicated that the east abutment wall has delaminated. In
addition, elastomeric bearing pads on this abutment presented excessive
deformation, and the bearing under beam 15 was cracked.
The unacceptably low load rating, determined by the consulting engineer, coupled
with the observations and conclusions provided in the field inspections and
monitoring program brought to light many of the structural deficiencies of the
westbound Sixth Avenue Viaduct. Additionally, these provided engineers with insight
as to which members needed repair as well as those which required further, more in
16


depth studies in order to determine appropriate actions.
Due to the substandard deck rating, testing was required of various members of the
deck in order to verify or revise the deck rating. Therefore, concrete core and
reinforcing bar samples were removed from the deck for further testing. A number of
tests were completed on the concrete core samples removed from the deck. The
primary testing supplied information about the compressive strength of the original
concrete deck; therefore, core compression testing was completed. In addition, the
consulting engineer wanted to prove that the white topping applied in the 1998
rehabilitation project was integral with the original deck. Consequently, a shear test
was required to determine the bond strength at this interface. While not required, the
modulus of elasticity was also tested as it provided additional valuable information
about the insitu condition of the original concrete in the bridge deck.
The primary purpose in testing the reinforcement bars was to determine their tensile
strength. Additionally, there were rare observations of areas where lower level
reinforcement bars were exposed and appeared corroded. Therefore, a representative
exposed reinforcement bar was tested to ensure that strength had not been
compromised. The totality of this testing completed on the bridge deck was then used
to revise the deck rating.
Over the years, inspectors have increasingly observed spalling of the concrete pier
columns, as shown in Figure 1.6. Also, according to a preliminary structural analysis,
each pier column lacks the flexural capacity to handle the demand loads. As a result,
the consulting engineer has recommended replacing each pier with a new reinforced
concrete pier.
17


Figure 1.6: Typical Concrete Spalling of Westbound 6th Avenue
Viaduct Pier Columns
However, according to the UPRR design criteria, 25 feet of horizontal clearance is
required from the piers of bridges to the centerline of the tracks. This requirement
applies to all new piers or existing piers that are to be rehabilitated. If this cannot be
provided, the piers must be protected by a crash wall or be built of heavy
construction. Heavy construction is defined by the UPRR as a pier having a cross
sectional area of at least 30 ft If heavy construction or a crash wall is provided, a
minimum of 9 feet of horizontal clearance must be provided between a given pier and
track centerlines. If these requirements cannot be met, a request for variance must be
obtained and approved by the UPRR. It is important to note that several UPRR tracks
were either permanently closed or realigned during the eastbound bridge
rehabilitation, thus creating additional horizontal clearance. Piers 13 and 17 are
unique in that they are positioned within the UPRR clearance requirements, and
therefore necessitated further testing of insitu elements in order to determine strength
18


parameters. Testing of these piers was aimed at providing engineering evidence to
either support leaving the current piers in place or to rebuild utilizing either a crash
wall or heavy construction so as to meet the UPRR requirements.
Pier 11, the fixed pier, needed to be further examined to determine if the existing pier
could safely hold required loads through the rehabilitation process, and to verify its
need for replacement. Therefore, a testing program incorporating various NDE testing
techniques was developed in order to determine the insitu condition of these piers.
During the preliminary design phase of the project, a test pit was excavated at the
west abutment to investigate the limits of the concrete spalling of the abutment cap.
The west abutment and abutment cap are shown in Figure 1.7. In the investigation of
the abutment cap, significant corrosion of the concrete-filled pipe piles was
discovered and additional test pits were excavated at the northernmost and
southernmost piles. Based upon these observations, City engineers wanted to
determine whether significant section loss had occurred thus compromising the
structural integrity of these members. As a result, testing was completed to verify the
thickness of the steel of the composite piles as compared with the plans, and to
determine the extent of section loss.
19


Figure 1.7: West Abutment of the Westbound 6th Avenue Viaduct
To determine if the structure could be rehabilitated to add an additional 20 to 30 years
of life, questions arose as to the structural integrity of various components of the
bridge; in particular, the deck, the pier columns, and the composite piles of the west
abutment, had to be tested.
As the bridge continues to provide an ever-increasing role in the overall
transportation scheme for the City of Denver, it was obvious that the only viable
means of testing these components would have to be non-destructive in nature, as the
viaduct would have to remain open through the duration of testing. Therefore, the
following scope and procedure for testing was devised so as to obtain the data
required concerning the various components of the bridge while minimizing impacts
to traffic as well as minimizing the destructive nature inherent in various forms of
bridge testing.
20


1.3 Scope of the Project
Engineers with the City and County of Denver concluded that it was necessary to
quantify the condition states of various components in order to assess the long-term
viability of the structure. The data obtained through this testing was necessary for
multiple reasons. Primarily, testing was required for deciding upon the necessity of
repair on an element level basis. Additionally, testing would provide evidence for the
verification or revision of load ratings and assumptions. Finally, test results would
also be utilized to determine the construction sequence of future repair project. It was
concluded that these needs could be readily met by applying a series of non-
destructive testing techniques to the individual components in question. The results
from this testing will provide an overall depiction of the condition of each of the
components.
The scope of this thesis is to:
Review previous literature and research on bridge testing. In particular, a
review of non-destructive testing techniques, and their application for
evaluating the condition of the various bridge components. This includes:
o Visual Inspection
o Hammer Sounding
o Rebound Test Hammer (Schmidt Hammer)
o Ultrasonic Pulse Velocity
o Tomography
o Concrete Core Sample Testing
Compressive Testing
Modulus of Elasticity Testing
Shear Testing
21


o Rebar T ensi le T esting
o Steel Thickness Ultrasonic Testing
Research coring procedures, applications and equipment
Research the use of strain gauges in the determination of modulus of elasticity
of concrete samples
Research UPRR Clearance Requirements
Research ASTM Testing Procedures
Develop a comprehensive non-destructive testing program to determine the
insitu condition of the specified components of the bridge
Complete the testing program
Determine the current condition of the components
Evaluate the long-term safety of the bridge based on the data gathered
22


1.4 Testing Procedure
Based on the scope and goals of the project, the following experimental procedure
was developed to provide the necessary information for determining the condition of
the particular components of the westbound 6th Avenue Viaduct.
1. Bridge Deck Testing
a. Steel Reinforcement Tensile Testing per ASTM A370 Annex A9
b. Compression Testing of Deck Core Samples
c. Modulus of Elasticity Testing of Deck Core Samples
d. Ultrasonic Pulse Velocity Direct Transmission Testing of Concrete
Deck Samples
e. Original Concrete Deck / White Topping Shear Testing
2. Pier Column Testing
a. Compression Testing of Pier Core Samples
b. Modulus of Elasticity Testing of Pier Core Samples
c. Ultrasonic Pulse Velocity Direct Transmission Testing of Pier Samples
d. Visual Inspection of Pier Condition
e. Hammer Sounding of each Pier
f. Schmidt Hammer Testing
g. Ultrasonic Pulse Velocity Direct Transmission Testing of each Pier
h. Horizontal Ultrasonic Tomograms
i. Vertical Ultrasonic Tomograms
3. West Abutment Testing
a. Composite Pile steel thickness testing using ultrasound techniques
The testing completed, as outlined in this procedure, provided researchers with
valuable information about the insitu condition of these components. This
23


information was then utilized to develop conclusions as to the repairs necessary in
order to extend the life of the bridge an additional 20-30 years. The totality of this
testing thereby provided engineers at the City and County of Denver with the
information necessary to move forward with repair efforts.
24


2. Previous Research
2.1 Bridge Testing
The majority of the 600,000 bridges in the U.S. highway system were built during
two periods of time, these being the 1930s and the 1950s 1960s. Approximately
30% of these 600,000 bridges are classified as being either functionally obsolete or
structurally deficient. [Hadavi, 1998; Tang, 2003] This classification does not mean
that the deficient bridges are necessarily unsafe; they are classified administratively to
indicate that they require some form of maintenance or major rehabilitation to restore
them to their original condition or to their current load carrying capacity. The average
age of the bridges in the US Interstate system is 45 years. [Tang, 2003] As a
consequence of the time periods of major bridge construction, significant portions of
the nations bridges are presently approaching a condition state that will necessitate
significant repairs, rehabilitation or even replacement.
On December 15, 1967, the collapse of the 2,235-foot long Point Pleasant Bridge,
over the Ohio River between West Virginia and Ohio, illustrated the need for
programs for the inspection and maintenance of bridges. [Hartle et al., 1990], The
Point Pleasant Bridge was built in 1928 and its failure occurred without warning,
resulting in the death of 46 people. Because of its deadly consequences, the collapse
exposed the necessity of a rational program to conduct periodic inspections of the
nations bridges.
Bridges can fail for several reasons: scour, wind, fatigue, earthquakes, floods,
corrosion, failure of a member, inappropriate design, and fire. [Taly, 1998, Haik et
al., 1990] Maintenance inspections and actions need to be periodically undertaken to
detect and prevent such failures.
25


The collapse of the Point Pleasant Bridge led to a national concern about the safety of
each bridge in the United States, and Congress was urged to create a national bridge
inspection standard. As a result, in 1971 the National Bridge Inspection Standard
(NBIS) was created to institute federal parameters for bridge inspections, report
formats, inspector qualifications, and inspection procedures. During the early 1970s,
manuals for bridge inspections were created and adopted. The Federal Highway
Administration (FHWA) Bridge Inspectors Training Manual 70 was first published
in 1970 and was used in training programs for bridge inspectors for several years.
Another important manual, the FHWA Recording and Coding Guide for the Structure
Inventory and Appraisal of the Nations Bridges was released in 1972. During this
time, it became clear that fund availability did not meet the maintenance costs
required for the bridge inventory. Therefore, in 1978 the Surface Transportation
Assistance Act created guidelines for funding maintenance and replacement of all
public bridges over 20 feet in length. The NBIS program had formerly restricted such
guidelines solely to bridges in the main federal highways. The Bridge Inspectors
Training Manual (BITM) 90 [Hartle et al., 1990] revised, upgraded, and replaced the
Bridge Inspectors Training Manual 70. This new manual is divided into 21 chapters
and brings detailed and comprehensive information about the inspection and
evaluation of bridge components and bridge inspection reporting systems. The BITM
90 also presents a chapter about the application of Non-Destructive Testing, referred
to as Advanced Inspection Techniques, in the inspection of bridges. In addition, the
FHWA initiated the Intermodal Surface Transportation Efficiency Act of 1991
(ISTEA). ISTEA mandated the creation of systems to manage bridges in each state
Department of Transportation and Metropolitan Planning Organization. It also funded
long-term research projects that address problems of the next century. In the later part
of the 1990s, new federally funded programs have been established, such as
NEXTA, BESTEA (Building Efficient Surface Transportation and Equity Act of
26


1997) , and ISTEA II. In addition to the activities referred in the ISTEA of 1991, these
new programs were aimed to address specific topics such as a timber bridge program,
scour counter measures, research of innovative materials, load and resistance factor
design specifications, and application of NDE in bridge assessment. [Densmore,
1998] The Federal Government spent $4 Billion per year through apportionments
during the 1998-2003 Transportation Equity Act of 21st Century legislation (TEA-
21). The states and local agencies spent about the same amount from their combined
matching shares and other tax revenues, thus doubling the annual spending to $7 $8
Billion for bridge improvements. The highway infrastructure continues to face
numerous challenges, i.e., increasing growth demands and heavier trucks as well as
trying to preserve aging and rapidly deteriorating highway bridges. The TEA-21
legislation launched an important initiative and established the Innovative Bridge
Research and Construction (IBRC) Program. The IBRC Program was one of the
largest Federal Government funded initiatives in the world; it was crafted in part to
seek effective means for extending the service life of the continually aging bridge
inventory.
27


2.2 Non-Destructive Testing Methodologies
To help deal with the problem of aging bridges, a number of methods have been
developed in recent years to help engineers address the problem of maintaining a
functional bridge network while limiting the number of bridge closures or weight
restrictions. Bridge management systems and deterioration modeling methodologies
have emerged as useful methods of predicting needs, prioritizing projects, and
directing funding. Unfortunately, these methods cannot provide empirical data on the
actual condition and behavior of deteriorated and distressed bridge infrastructure.
Often such data are required if critical decisions regarding safety, prioritization, and
feasibility can be made. As a result, field bridge testing has become increasingly
popular as a tool for engineers to assess the performance of bridges. [Saraf and
Nowak, 1998; Gogel, 2001] Such testing provides valuable information for
prioritizing repairs and rehabilitation projects on both a case specific and a network
level. Indeed, many agencies are adopting network-wide testing programs for
evaluation of bridge infrastructure. [Moses et al, 1994; DeWolf et al 1998] In
addition to addressing the specific needs within an agency, increased use of bridge
testing under service level conditions has led to a better understating of actual
structural behavior.
A growing need for data on the insitu performance of bridges is currently being met
by growing interest and practice in the area of field bridge testing. Currently, field-
testing of bridges serves two principal purposes:
1. Verification of models and assumptions used in bridge design
2. Evaluation of existing infrastructure [Ryall, 2001]
28


Testing for both purposes has become more popular as improved technology and
declining equipment costs in recent years have made bridge testing more feasible than
ever.
As bridge testing has increased in recent years, so have the number of applications of
testing. [Bakht and Jaeger, 1990; DeWolf et al, 1998; Chajes et al, 1997, Boothby and
Craig, 1997; Mohammadi et al, 1998; Saraf and Nowak, 1998; Sartor et al, 1999;
McElwain and Laman, 2000; Prine, 1996; Moorty and Roeder, 1992; Moses et al,
1994; Farhey, 2000]
The scope and duration of bridge testing is generally tailored to the specific goals of
the study. In some cases a bridge is tested in a single day, while other studies may
involve permanent instrumentation or retesting carried out at specified intervals.
Often specific elements of a bridge are targeted for instrumentation because initial
analysis or visual inspection suggests areas of concern. While some of the specific
methodology being used for various applications of field-testing is relatively new,
much of the technology being used in these applications has been in use for several
decades. For the most part, tests are aimed at the assessment of strength and other
fundamental material properties or to locate and obtain comparative results indicating
permeable regions, cracks, laminations, and areas of lower integrity than the rest.
Coupled with detailed visual investigations, the diagnostic non-destructive testing
results can assist the engineer to decide upon the appropriate rehabilitation steps. One
can compare such diagnostic tests aimed at identifying the defects in the structure to
clinical tests aimed at diagnosing the diseases in the body. Such testing enables the
engineer (doctor) to plan and administer the appropriate medication (the repairs and
strengthening) so as to cure the structure (patient) as best as possible within the
prevailing constraints of ability and funds.
29


Once a bridge has been inspected for its overall visual condition, it is often necessary
to carry out non-destructive tests in order to further extend the diagnostic process if it
is suspected that the bridge has been weakened in some way. Non-destructive in this
context means that the structure is not destroyed, though parts of it may be broken
away for testing or inspection.
The tests are normally ordered to determine:
The physical quality of the materials
The position and extent of hidden defects, elements and material boundaries
The tests are carried out insitu and provide further information from which an
improved diagnosis can be made to enable the bridge engineer to make decisions on
the necessary remedial work. Some of the tests are carried out during the course of a
normal inspection while others are only applicable during a special inspection.
The importance of applying more powerful and accurate tools as a supplement of the
visual inspection has been addressed in the BITM. [Hartle et al., 1990] The manual
refers to NDE techniques in Chapter 15 Advanced Inspection Techniques. Two of
the applications of NDE listed in the BITM are:
Evaluation of defects found in the visual inspections
Inspection of components that cannot be readily evaluated using visual
inspection
NDE techniques have many advantages when compared with other conventional
inspection techniques. The most obvious of these advantages being that there is no
damage to the structure as a result of the testing, and that NDE techniques can be
30


applied during in-service inspections. Additionally, NDE techniques allow for the
detection of flaws and defects throughout the member, including internal defects that
may otherwise go undetected. Depending upon the method, a very precise
characterization of the defect extent can be achieved. Thus, NDE techniques can
provide a very detailed glimpse into the overall condition of a member. Nevertheless,
because of the inherent complexity there are several obstacles to the application of
NDE techniques in bridge management systems. The primary obstacle being that
NDE equipment requires experience and expertise in the operation of equipment as
well as in the interpretation of the results. The results from NDE testing are complex,
providing very detailed quantitative information, thus requiring expertise in order to
interpret the results. Another important factor in the use of NDE is the interpretation
of the results. Misleading interpretations can produce disastrous conclusions in bridge
management administration.
The application of NDE methods must be incorporate visual inspections and be
restricted to a limited subset of structures in the entirety of the bridge network. The
techniques should be implemented to address specific problems in the bridges and to
obtain additional information of deterioration processes already identified or
suspected to exist by the inspector. Although there is a fundamental difference
between the qualitative results from a visual inspection and the quantitative output by
applying NDE methods, the visual condition ratings can be used to help assess the
severity of the deterioration process. Therefore, the separation between the visual and
NDE inspections cannot be ignored in any bridge inspection program.
Prior to the application of an NDE method, the most adequate technology for the
problem needs to be determined. The type, severity, and extension of the
deterioration are parameters that help in the selection of the most appropriate NDE
method in each deteriorated bridge.
31


2.2.1 Visual Inspection
In the visual inspection of a bridge, the inspector carries out a subjective operation by
noting the severity and multitude of various defects. In particular, the inspector
examines the structure as a whole or the particular member noting defects such as
cracks, spalls, corrosion, section loss, and any other visible defects in the structure. In
this operation, the past experience of the bridge inspector plays an important role. The
subjective decision in the condition rating assignment may deliver a good indicator of
the general condition of the bridge but cannot be the only parameter for decisions on
maintenance actions.
32


2.2.2 Hammer Sounding
The hammer sounding method is a sonic non-destructive testing method used to
determine the condition of the concrete beneath the surface. The test is performed by
striking the concrete surface with a common carpenters hammer and carefully
listening to the impact sound. A crisp sound indicates that the concrete is solid and is
recorded as good. A muffled thud sound indicates voids, delaminations, cracks, or
some other deficiency and is recorded as bad. In order to graph the results of the
qualitative hammer-sounding test, a numeric scale must be devised to quantify the
results.
For the purposes of this study the Golden Software Surfer package was utilized to
display contour plots of the hammer sounding data.
33


2.2.3 Rebound Test Hammer
A rebound test hammer, commonly called a schmidt hammer is an instrument used
for the non-destructive testing of hardened concrete. Whereby, the rebound of a steel
hammer from the concrete surface indicates the hardness of the concrete, which can
then be related to its strength. Thus, this method provides a means for field
measurement of one of the primary measures of the quality of concrete its
compressive strength.
The schmidt hammer is simple to use, very portable and cost efficient. Its use may be
easily employed to identify areas to be further tested with alternate NDE methods.
The hammer itself is a self-contained unit and is shown in Figure 2.1. The tip of the
unit, the impact plunger, is held vertically or horizontally perpendicular to the smooth
surface of the concrete and is pushed. This compresses an internal spring that
automatically lifts and releases a hammer mass on the impact plunger and then to the
concrete. The impact energy is well defined and the rebound of the hammer mass is
dependant on the hardness of the concrete. Rebound values are indicated on a gauge
built into the instrument. It is mostly a relative or comparative technique since the
absolute value depends on the local variations in the surface properties due to the
presence of voids or aggregate particles. A number of measurements are therefore
required in the same location from which the mean and standard deviation values can
be determined. Since the calibration of the most well known rebound hammers is
based on conventional testing on 200 mm cubes, it is necessary to multiply hardness
values by a form factor to take account of different sized specimens.
34


Figure 2.1: Rebound Test Hammer
Great care is necessary when using the compressive strength derived in this manner
for use in strength assessment of bridges because the values are limited to the surface
of the concrete only and cannot automatically be extrapolated to the body of the
concrete. Although surface hardness methods are well established and relatively
inexpensive they have limited application in assessing distressed structures. ASTM
standard C 805-85 discusses the methodology for the application of the rebound test
hammer methods. Since this method only evaluates the surface of the concrete it has
limited use in massive structures. Furthermore, this method is considered usable only
in relatively new structures. [Minor et al., 1988] For the use in old concrete, direct
correlation with compressive strength of cores is necessary. [Malhotra and Carino,
1991] While its use for determining compressive strength may be limited to relatively
newer structures, results from older structures can provide comparative results of the
surface hardness over the area tested.
For the purposes of this study the Golden Software Surfer package was utilized to
display contour plots of the average rebound number data.
35


2.2.4 Ultrasonic Pulse Velocity
The Ultrasonic Pulse Velocity (UPV) method provides a considerable amount of
information about the interior condition of a plain, reinforced, or prestressed concrete
member. The basic principle is that the measurement of the velocity of an ultrasonic
pulse transmitted through the concrete specimen can assess the quality of concrete as
the velocity of sound waves traveling through a medium depends upon the elastic
properties of that medium. Thus, by knowing the pulse travel time and the geometry
of the specimen being tested, the velocity can be determined which allows for the
estimation of the specimens elastic properties. UPV testing can also be utilized in
the detection of internal defects as the pulse travel time is increased where regions of
voids, cracking, honeycombing and other imperfections are encountered. Correlation
with control specimens results in a relative picture of the specimen.
Modem ultrasonic equipment is lightweight and portable; simple to operate and has a
high order of accuracy and stability. The instmment used for the purposes of this
study is shown in Figure 2.2. The testing equipment is comprised of a transmitting
transducer that generates an impulse, which is applied to the concrete surface. A
similar receiving transducer is located at a known distance on the concrete surface
opposite of the transmitting transducer (for the direct transmission method). Both
transducers are coupled to the concrete such that air does not exist between the
concrete and transducer surfaces. The electronic timing device measures the pulse
travel time. The pulse velocity measurement depends on accurate measurement of the
path length and the transmit time of the pulse. This demands that there is good contact
between the flat surface of the metal transducer and the concrete. A coupling agent is
often used to ensure good contact between the transducers and the specimen and thus,
good pulse transmittance.
36


Figure 2.2: Ultrasonic Pulse Velocity Testing
The information gathered can help to provide a quick diagnostic picture for the bridge
engineer and can enable him or her to plan a more focused testing program if
necessary. Bungey and Millard (1996) describe the method very thoroughly and
although they give detailed information on how to apply the method, they do urge
caution in the interpretation of results.
The pulse velocity (V) of the longitudinal ultrasonic vibrations traveling in an elastic
medium is given by:
V =
E(l-v)
p( \ + v)(\~2v)
-il/2
(2.1)
where: E = dynamic elastic modulus
p = density
v = Poissons ratio of the medium
37


Typical correlation of pulse velocity with quality is shown in Table 2.1.
Table 2.1: Correlation of Pulse Velocity with Concrete Quality [Ryall, 2001]
Longitudinal Pulse Velocity (km/sec) Quality of Concrete
>4.5 Excellent
3.5-4.5 Good
3.0-3.5 Doubtful
2.0-3.0 Poor
<2.0 Very Poor
Pulse velocity measurement can be made in three modes as indicated in Figure 2.3,
namely:
1. Direct Transmission
2. Semidirect Transmission
3. Indirect or Surface Transmission
Figure 2.3: Ultrasonic Pulse Velocity Transmission Modes (a) direct; (b) semi-direct;
(c) indirect
38


The instrument measures the time for the pulse to travel from the emitter to the
receiver. Where opposite faces of a member are exposed such as a slab, beam or
column, the direct mode is used and is preferred as the path is well defined. This is
also the favored transmission mode to be utilized for void detection. In some cases all
faces of bridge elements are not accessible and the indirect modes are used. The path
is less easily defined but it is the ideal arrangement when measuring the presence and
depth of surface cracks.
An ASTM standard (C 597-83) to use pulse velocity in concrete brings information
about the application of this method. The measurements of the pulse speed can be
used to determine the quality of the concrete compressive strength. [Krautkramer and
Krautkramer, 1990]
A newly developed approach to the application of ultrasound in structural inspections
is called Direct-Sequence Spread-Spectrum Ultrasonic Evaluation (DSSSUE) and has
been tested in bridge components. [Wormley et al., 1995; Rens et al., 1997] Although
this technique has been successfully used it is not appropriate to be applied in
localized problems. This is because it performs a global analysis of the structure,
which detects several changes in the system properties at once. Furthermore,
permanently mounted transducers may be required to prevent variations due to
couplant changes. [Wormley et al., 1995]
For the purposes of this study the Golden Software Surfer package was utilized to
display contour plots of the ultrasonic pulse velocity direct transmission data.
39


2.2.5 Tomography
Acoustic tomography uses UPV methods to define an entire interior plane within a
concrete member. While the UPV method alone provides information regarding an
abnormality beneath the surface, a tomogram can provide an actual cross-sectional
image beneath that same surface. Sound waves are transmitted through the surface
from an array of source and receiver combinations. Travel times and geometry data
are processed with the aid of a tomography software program and the resulting
product is an image of the interior plane.
It is possible to use the same equipment that is used in ultrasonic testing to perform
tomographic analysis of concrete. [Chang and Wang, 1997; Transue et al., 1997,
Liephart et al 1999] The tomography software computer program uses a large number
of pulse velocity readings obtained on the exterior of the specimen and delivers the
map of velocities in the interior through the application of reconstruction algorithms.
[Schuller and Atkinson, 1995] Gamma ray tomography can accurately detect many
different phases of reinforced concrete deterioration but the applicability of this
method on large structures has not yet been determined. [Martz et al., 1994] Although
acoustic tomography is not as accurate as X-ray or gamma ray systems the costs are
comparatively small.
For the purposes of this study, the GeoTomCG software package was utilized to
analyze the tomography data and to display the results.
40


2.2.6 Concrete Core Specimen Removal
The most accepted and standardized method of determining the insitu strength of
concrete is the laboratory testing of concrete cores drilled from the structure in
question. Cores of concrete from the structure are drilled with the help of a core drill
utilizing diamond tipped bits. Cores of concrete are very often extracted from insitu
structures in order to be analyzed in the laboratory to determine such properties as
density, tensile strength, compressive strength, carbonation, and permeability. The
diameter of the core should be as large as is practically possible to ensure that the
local effects of aggregates do not adversely affect the results. A water-cooled
diamond-tipped coring drill bit is attached to the drill, which is mounted on a stand
and can be bolted to the surface of the concrete being examined. A typical
arrangement is shown in Figure 2.4.
Figure 2.4: Concrete Core Drilling Equipment
If the permeability of the core is not being determined, then after the core has been
extracted, it is trimmed at each end using a water-cooled diamond tipped rotary saw.
Care should be taken not to cut through any steel reinforcement, or if this cannot be
avoided, then a judgment must be made as to whether the presence of the
41


reinforcement in the core sample might affect the results of testing to be performed on
the core sample. More importantly, an engineer must decide whether such local
damage might affect the overall strength of the bridge. With insitu concrete in
existing bridges only a relatively small number of cores can be taken, and these must
be chosen with care to be representative of the area being tested. Concrete core
samples obtained from insitu bridge components can provide numerous forms of
valuable information about the actual physical condition of the member in its
environment at the time of specimen removal.
42


2.2.7 Core Compression Testing
Compression testing of concrete is carried out in a suitable compression testing
machine at a controlled speed. The ends can either be trimmed square with a disc
cutting saw or leveled using a suitable epoxy resin compound.
Stress is defined as the object's internal resisting forces. For a uniform distribution of
resisting forces, stress can be calculated by dividing the force (F) applied by the unit
area (A):
Stress (cr) = F/A (2.2)
Cubes and cylinders of the same concrete have different crushing strengths; the
cylinder falling at about 80% of the cube strength. [Ryall, 2001] The cylinder strength
probably gives a more realistic estimate of the uniaxial strength of the concrete as it
fails in shear at a natural angle; whereas, the cube is restrained by the platens on the
testing machine such that it is in a confined space. A typical cylinder test set-up is
shown in Figure 2.5.
43


Figure 2.5: Compression Testing Machine
In performing the core compression test, the ends of the specimen must first be
trimmed square, and accurate measurements of the cross-sectional area must be taken
prior to testing. The specimen is then placed in the center of the compression testing
machine. Once the specimen is in place, load is slowly applied to the specimen. At
some point during the application of load, the specimen will begin to crack, and the
compressive load will be at a maximum just prior to cracking. This maximum
compressive force is then divided by the cross sectional area to obtain the ultimate
compressive stress of the specimen.
44


2.2.8 Core Modulus of Elasticity Testing
The modulus of elasticity of concrete core samples can be determined by utilizing
strain gauges while performing the core compression test. Unlike the core
compression test, load is typically applied until reaching approximately 40% of the
ultimate compressive strength.
When external forces are applied to a stationary object, stress and strain are the result.
Strain is defined as the displacement and deformation that occurs as a result of the
application of load. More specifically, strain is defined as the amount of deformation
per unit length of an object when a load is applied. Strain is calculated by dividing the
total deformation by the original length (L):
Strain(f) = (AL)/L (2.3)
Typical values for strain are less than 0.005 inch/inch and are often expressed in
micro-strain units:
Micro strain (jus) = Strain (£) x 10'6 (2.4)
Strain may be compressive or tensile and is typically measured by strain gages.
The modulus of elasticity (also known as the elastic modulus, youngs modulus and
stress-strain ratio) is the ratio of stress (nominal) to corresponding strain below the
proportional limit of a material. It is the slope of the straight-line portion of a stress-
strain diagram, and is expressed in units of force per area. The modulus of
elasticity may therefore be determined by dynamic testing whereby measurements of
45


stress (load) and strain are simultaneously observed. Strain gauges are a proven
technology capable of measuring differential strain in a specimen.
In performing the modulus of elasticity testing, appropriate strain gauges are affixed
to the specimen as outlined by the manufacturer. The strain gauges are then connected
to a strain indicator, as shown in Figure 2.6, which will display the strain in the gauge
throughout the testing.
The core specimen is then placed into the compressive testing machine. Load is
slowly applied as measurements of stress (load) and strain are simultaneously
observed. Load is typically applied until reaching approximately 40% of the ultimate
compressive strength. For each specimen, 3-4 such loading series should be used to
arrive at a stress-strain curve and average modulus of elasticity.
The deformation of an object can be measured by mechanical, optical, acoustical,
Figure 2.6: The P-3500 Digital Strain Indicator (left)
and SB-10 Switch and Balance Unit (right)
46


pneumatic, and electrical means. The most widely used characteristic that varies in
proportion to strain is electrical resistance. Although capacitance and inductance
based strain gages have been constructed, these devices' sensitivity to vibration, their
mounting requirements, and circuit complexity have limited their application.
Fundamentally, all strain gages are designed to convert mechanical motion into an
electronic signal. A change in capacitance, inductance, or resistance is proportional to
the strain experienced by the sensor. If a wire is held under tension, it gets slightly
longer and its cross-sectional area is reduced. This changes its resistance in
proportion to the strain sensitivity of the wire's resistance.
The ideal strain gage would change resistance only due to the deformations of the
surface to which the sensor is attached. However, in real applications, temperature,
material properties, the adhesive that bonds the gage to the surface, and the stability
of the metal all affect the detected resistance. Because most materials do not have the
same properties in all directions, knowledge of the axial strain alone is insufficient for
a complete analysis.
When selecting a strain gage, one must consider not only the strain characteristics of
the sensor, but also its stability and temperature sensitivity. Unfortunately, the most
desirable strain gage materials are also sensitive to temperature variations and tend to
change resistance as they age. For tests of short duration, this may not be a serious
concern, but for continuous industrial measurement, one must include temperature
and drift compensation. Each strain gage wire material has its characteristic gage
47


factor, resistance, temperature coefficient of gage factor, thermal coefficient of
resistivity, and stability.
The ideal strain gage is small in size and mass, low in cost, easily attached, and
highly sensitive to strain but insensitive to ambient or process temperature variations.
Figure 2.7: Typical Strain Gauges
In bonding strain gage elements to a strained surface it is important that the gage
experience the same strain as the object. With an adhesive material inserted between
the sensors and the strained surface, the installation is sensitive to creep due to
degradation of the bond, temperature influences, and hysteresis caused by
thermoelastic strain. Because many glues and epoxy resins are prone to creep, it is
important to use resins designed specifically for strain gages.
In order to measure strain with a bonded resistance strain gage, it must be connected
to an electric circuit that is capable of measuring the minute changes in resistance
corresponding to strain. Strain gage transducers usually employ four gage elements
electrically connected to form a Wheatstone bridge circuit.
48


A Wheatstone bridge, as shown in Figure 2.8, is a divided bridge circuit used for the
measurement of static or dynamic electrical resistance. The output voltage of the
Wheatstone bridge is expressed in millivolts output per volt input. The Wheatstone
circuit is also well suited for temperature compensation. [Horrowitz & Hill, 1989]
< 1 Output mV C
V|N j B ^ VOUT D N A
^sv-si

Figure 2.8: Wheatstone Bridge Schematic
In Figure 2.8, if Rl, R2, R3, and Rg are equal, and a voltage, VIN, is applied between
points A and C, then the output between points B and D will show no potential
difference. However, if Rg is changed to some value that does not equal Rl, R2, and
R3, the bridge will become unbalanced and a voltage will exist at the output
terminals. In a so-called G-bridge configuration, the variable strain sensor has
resistance Rg, while the other arms are fixed value resistors.
The strain gauge sensor, however, can occupy one, two, or four arms of the bridge,
depending on the application. The total strain, or output voltage of the circuit (Vout)
is equivalent to the difference between the voltage drop across Rl and Rg. The bridge
is considered balanced when R1/R2 = Rg/R3 and, therefore, Vout equals zero. Any
49


small change in the resistance of the sensing grid will throw the bridge out of balance,
making it suitable for the detection of strain. When the bridge is set up so that Rg is
the only active strain gage, a small change in Rg will result in an output voltage from
the bridge. This output voltage indicates the strain experienced by the gauge.
The output of a strain gage circuit is a very low-level voltage signal requiring a
sensitivity of 100 microvolts or better. The low level of the signal makes it
particularly susceptible to unwanted noise from other electrical devices. Capacitive
coupling caused by the lead wires running too close to AC power cables or ground
currents are potential error sources in strain measurement. Other error sources may
include magnetically induced voltages when the lead wires pass through variable
magnetic fields, parasitic (unwanted) contact resistances of lead wires, insulation
failure, and thermocouple effects at the junction of dissimilar metals. The sum of such
interferences can result in significant signal degradation. [Horrowitz & Hill, 1989]
50


2.2.8.1 Strain Gauge Installation Procedures
The materials necessary for strain gauge installation are shown below, in Figure 2.9.
Figure 2.9: Materials for Gauge Preparation and Installation
Work Surface Preparation
1. Find a suitable work area and wipe it clean
2. Mount the metal plate onto the work surface
3. Clean the metal plate with degreaser
Lead Wire Preparation
4. Cut lead wire into uniform lengths
5. Separate the red, black, and white sections of the wire and strip lA of
insulation from the end of each wire
6. Separate and strip Vi" of insulation from the other end of each wire
51


7. Twist the wire ends together: black/white and red
8. Tin both ends of each lead wire
Gauge Solder Termination Preparation
9. Clean forceps with Neutralizer 5A
10. Remove the Strain Gage and place it in the center of the cleaned plate with
soldering tabs face up
11. Place a 1-2 piece of masking or drafting tape across the gauge leaving approx
half of the soldering tabs exposed
12. Tin the soldering tabs on the strain gage
Soldering the Lead Wires to the Gauges
13. Lay the lead wire on the plate in position
14. Tape the lead wire in place
15. Using tweezers, bend the lead wire to the correct positions
16. Solder the wire to the tabs
Gauge Removal and Cleaning
17. Brush several applications of rosin solvent over the gauge, tape and solder
joint
18. When the solvent has loosened the tape from the gauge, lift the gauge and dip
into a small container of clean rosin solvent
19. Blot the gauge dry
20. Wipe the gauge surfaces with Neutralizer 5A
21. Allow to air dry for approximately 5 minutes
22. Store strain gauge
52


Concrete Core Preparation
Degreasing
23. Spray CSM-1A Aerosol degreaser in the appropriate areas
24. Wipe from top to bottom
25. Repeat the degreasing process a second time
Abrading
26. Use a wire brush and file to abrade the surface, while keeping the surface wet
with M-Prep Conditioner
27. Use sand paper to abrade, keeping wet with conditioner
Placement of Gauge Layout Lines
28. Mark the specimen at appropriate locations for strain gage placement
Conditioning
29. Wash the test surface with a generous amount of Conditioner A and cotton
swabs (wiping in one direction until cotton is no longer discolored)
30. Wipe with a gauze sponge
31. Repeat in opposite direction
Neutralizing
32. Apply liberal amounts of Neutralizer 5 A and wipe in one direction
33. Wipe with gauze sponge
34. Repeat in opposite direction
Check the functionality of the strain gauges
35. Measure the base resistance of the unstrained strain gage after it is mounted,
but before wiring is connected
53


36. Check for surface contamination by measuring the isolation resistance
between the gage grid and the stressed force detector specimen using an
ohmmeter, if the specimen is conductive. This should be done before
connecting the lead wires to the instrumentation. If the isolation resistance is
under 500 megaohms, contamination is likely.
37. Check for extraneous induced voltages in the circuit by reading the voltage
when the power supply to the bridge is disconnected. Bridge output voltage
readings for each strain-gage channel should be nearly zero.
38. Connect the excitation power supply to the bridge and ensure both the correct
voltage level and its stability.
Strain gauge installation
39. Clean a glass or metal plate with degreaser, conditioner and neutralizer
40. Place a strain gage on the plate with the bonding side down
41. Place a 4-6 piece of scotch tape over the gauge
42. Lift the strain gauge using the tape, and place on the specimen at the
appropriate location
43. Lift the tape to expose the strain gauge, and place glue on strain gauge
44. Press down the tape, and run finger (in gauze) over the gauge/tape to remove
air bubbles
45. Place a Teflon sheet over the gage, and a silicone rubber pad over the Teflon
sheet
46. Apply clamping pressure, and allow to set for 12-16 minutes
For a more thorough application procedure refer to Allen, 2002.
54


2.2.9 Core Shear Testing
In order to determine the bond strength at the interface of two materials, the double
shear test may be performed. In conducting the test, the specimen is carefully placed
into a double shear test block such that the interface lines up with one of the edges of
the block. An epoxy is placed along the rim of the test block to keep the specimen in
place through the testing process. The test block with specimen in place is then placed
into the compression testing machine and load is slowly applied, until the specimen
shears. In order to determine the bond strength at the material interface, the specimen
must shear at the interface location. However, if the specimen were to shear in the
body of a layer as opposed to at the interface, it would indicate that the bond strength
at the interface is greater than the shear strength through the specimen, and therefore,
the strength of the bond is not of greatest concern, but rather the shear strength of the
material is of greater concern. Assuming the specimen shears at the interface, the
bond strength at this interface can then be determined through the free-body diagram
and calculations of Figures 2.10 to 2.12.
55


Figure 2.10: Double Shear Test Block
Figure 2.11: Double Shear Test
56


/fiRAC.ll'Jfti- T>C< K
' I mi Cp-FRCC
Figure 2.12: Force Diagram of Double Shear Test
Where:
_V*Qna _V*(7rr2 l2){4rlZn) ^ AV
Tmax Tm I*tNA (nr4 / 4)(2r) 3nr2 3 A
Therefore, by slowly applying compressive load to a specimen placed in the double
shear test block, one can determine the maximum bond shear strength at an interface.
57


2.2.10 Rebar Tensile Testing
The tensile test is a method for determining behavior of materials under axial tensile
loading. Data from testing are used to determine elastic limit, elongation, modulus of
elasticity, proportional limit, reduction in area, tensile strength, yield point, yield
strength, and other tensile properties. Tensile tests at elevated temperatures provide
creep data. Procedures for tensile tests of metals are given in ASTM E-8. The tensile
strength is the ultimate strength of a material subjected to tensile loading, and the
tensile test determines the maximum stress that can be placed upon the material.
In conducting tensile testing on rebar samples, both ends of the sample are fixed into
the tensile testing device. Tensile load is slowly applied to the specimen. The point in
time when the change in deformation is no longer proportional to the stress, the
elastic limit is recorded. After passing the elastic limit, tensile loading continues until
the specimen fails.
A testing machine is used to strain the specimen and to measure the load required to
produce the deformation. The stress is obtained by dividing the load by the initial
cross-sectional area of the specimen. The cross-sectional area will change somewhat
during the loading.
The yield strength is defined as the stress that will induce a specified permanent set,
usually 0.05 to 0.3 percent, with 0.2 percent the most commonly used value. The
yield strength can be conveniently determined from a stress-strain diagram by laying
off the specified offset (permanent set) on the strain axis and drawing a line parallel
58


to the elastic portion of the graph. The stress indicated by the intersection of this line
and the stress-strain diagram is the yield strength for the specified offset.
The maximum stress developed in a material before rupture is called the ultimate
strength of the material, and the term may be modified as the ultimate tensile,
compressive, or shearing strength of the material. Ductile materials undergo
considerable plastic tensile or shearing deformation before rupture. When the ultimate
strength of a ductile material is reached, the cross-sectional area of the test specimen
starts to decrease or neck down, and the resultant load can be carried by the specimen
decreases. Thus, the stress based on the original area decreases beyond the ultimate
strength of the material although the true stress continues to increase until rupture.
59


2.2.11 Steel Thickness Testing
Steel ultrasonic thickness testing is a non-destructive test method whereby high
frequency sound waves are introduced into a test object to gain information about the
thickness of that object. In addition to the thickness, steel ultrasonic testing can
provide information as to the location of a discontinuity or flaw, as well as the size
and type of discontinuity or flaw.
The thickness of each member is an important property used to determine stress in
each member of the structure. UCD owns a Panametrics Epoch III Flaw Detector
(Epoch III) as shown in Figure 2.13. The Epoch III is a hand-held, portable ultrasonic
device used to measure the thickness of steel as well as to locate any flaws in a steel
sample (i.e., cracks, pits, porosity, etc.). The decrease in thickness of a member is
primarily caused by corrosion, and corrosion is normally caused by moisture.
60


Figure 2.13: Epoch III Ultrasonic Flaw Detector
Figure 2.14: Steel Ultrasonic Testing Display
61


The Epoch III Flaw Detector measures two separate quantities. The first of these is
the time for the sound pulse to travel to a discontinuity and back; this is referenced on
the horizontal axis on the flaw detector screen. The second measured quantity is the
amount of energy in the returning sound pulse and is referenced on the vertical scale.
From these two measurements all other information can be calculated.
The Epoch III can operate in any one of four different types of waveform rectification
modes (full wave rectify, half wave positive, half wave negative, or unrectified). The
Epoch III utilizes a shock excitation pulser that is designated to provide a
combination of good near surface resolution for testing thin materials and excellent
penetration power for testing thick material. There are three different energy settings
that can be selected depending on test conditions.
The thickness of a material sample will help to determine if the material has
undergone any section loss due to corrosion or surface wearing. By scanning a
material, the location of a discontinuity or flaw can be found. Once the location is
found, the technician can investigate the flaw more thoroughly to determine flaw size
and type.
For further information on the operation of the Epoch III Flaw Detector refer to
Gogel, 2001 or the manufacturers instruction manual.
62


3. Testing of the Deck
Due to the substandard deck rating, as previously discussed, testing was required of
various members of the deck in order to verify or revise the deck rating. Utilizing the
as-built specifications the concrete deck was rated 21.6 tons at inventory rating and
36.1 tons at operating level. The desired inventory rating is 36 tons. Therefore,
concrete core samples and reinforcing bar samples were removed from the deck for
further testing. A number of tests were completed on the concrete core samples
removed from the deck. The primary testing required information about the
compressive strength of the original concrete deck; therefore, core compression
testing was completed. In addition, in order to determine if composite action exists
engineers needed proof that the white topping applied in the 1998 rehabilitation
project was integral with the original concrete deck. While not required, the modulus
of elasticity was also tested as it provided additional valuable information about the
insitu condition of the original concrete in the bridge deck. The primary purpose in
testing the reinforcement bars was to determine their tensile strength. There were rare
observations of areas where lower level reinforcement bars were exposed and
appeared corroded. Therefore, a representative exposed reinforcement bar was tested
to ensure that strength had not been compromised. The totality of this testing
completed on the bridge deck was then used to revise the deck rating.
63


3.1 Deck Steel Reinforcement Tensile Testing
Lower level transverse reinforcement bars were obtained from three locations.
Sample #1 was chosen as it was clearly exposed on the underside of the bridge deck
and was weathered and corroded as shown in Figures 3.1 and 3.2. Testing of this
rebar was aimed at determining whether corrosion had compromised the tensile
capacity of the exposed and corroded rebars.
Figure 3.1: Location of Rebar Sample #1 (Exposed & Corroded)
Figure 3.2: Rebar Sample #1
Samples #2 and #3 were obtained from the shoulder of the bridge near Pier 17, as
shown in Figure 3.3. These samples were not exposed or corroded and were removed
using a concrete chipper and a grinding wheel. These samples were chosen for two
64


reasons; first, to determine the tensile strength of the lower level transverse
reinforcement bars, and secondly, to compare with the results from Sample #1 to
determine if the exposed and corroded rebars have a lower tensile capacity.
The results from the ASTM A370 Annex A9 tensile testing are listed below, in Table
3.1.
Table 3.1: Deck Lower Level Transverse Steel Reinforcement Bars Tensile Test
Results
Yield Strength Tensile Strength Elongation
Identity Load (lb) lb / in2 Load (lb) lb / in2 in / in %
#1 15,6501b 50,500 25,500 lb 82,000 9.1 14%
#2 16,550 lb 53,500 26,325 lb 85,000 8.92 11.50%
#3 17,450 lb 56,500 27,400 lb 88,500 9.22 15%
65


Testing experimentally determined the average yield strength of the lower level
transverse reinforcement samples to be 53,500 psi, and the average tensile strength
was 85,166 psi. The testing showed that both the yield and tensile strength of the
exposed and corroded sample (Sample #1) was not significantly lower than the other
samples. Thus, the data did not conclusively prove strength reduction of the exposed
and corroded reinforcement bars. Conservatively, it was decided that the lower of the
two yield strengths from the non-corroded samples would be used to revise the deck
rating. Thus, a yield strength of 53,500 psi was used. A comparison with the as-built
plans, which specify an allowable stress of 20,000 psi, revealed that the steel
reinforcement was providing ample strength.
66


3.2 Deck Concrete Core Compression Testing
Deck core samples were removed from the northern edge of the bridge near pier 17,
as shown in Figure 3.4. Special precautions had to be taken to ensure that the water
used during the coring process did not get into the motor of the core drill, as overhead
drilling was performed to remove the vast majority of the core samples.
Three of the core samples of the deck were tested to determine the compressive
strength of the original concrete deck layer. The results of this testing are shown
below in Table 3.2.
67


Table 3.2: Original Concrete Deck Compression Testing Results
Cylinder ID ULTIMATE Load Strength Adjusted Strength (lbs) (psi) (psi)
D2A 39800 6030 5810
D3A 35900 5430 5380
D6 44500 6750 6520
The adjustment factor alters the compressive strength if the specimen is not of an
adequate length to diameter ratio. The experimentally determined average of the
ultimate adjusted compressive strength of the three original deck core samples was
determined to be 5903 psi. This value is just above the specified ultimate compressive
strength requirement of 5000 psi for the original deck of the bridge.
It was decided that the lowest of the three ultimate adjusted compressive strengths be
used to revise the deck rating. Thus, a compressive strength of 5,380 psi was used.
68


3.3 Deck Concrete Core
Modulus of Elasticity Testing
One of the deck core samples (Deck Core D6) was mounted with four strain gauges
in order to determine the modulus of elasticity of the original concrete deck. The
strain gauges were mounted at the quarter points along the circumference of the core,
such that gauge 1 was opposite gauge 4, and gauge 2 was opposite gauge 3, as shown
in Figure 3.5. The core was loaded to approximately 40% of capacity, as
measurements of load and strain were simultaneously observed. Measurements were
taken over three loading periods.
Using the data collected from all four of the strain gauges, a stress strain diagram was
obtained, and a modulus of elasticity of 3.359 xlOA6 psi was determined.
69


3.4 Deck Core Shear Testing of
White Topping Layer
The strength of the bond between the original bridge deck surface and the white
topping was tested. Only two of the fourteen core samples removed had a bonded
deck and white topping. The other 12 cylinders were not able to be removed with the
bond still in tact, and therefore, were not capable of being used for this testing. Many
of these cylinders simply came out in two pieces, as the interface bond surface
seemed negligible or non-existent. Some of the bonds may have been inadvertently
broken either during the coring process or specimen removal.
Figure 3.6: Deck Core Layers
By completing a double shear test, as shown in Figures 2.10 to 2.12, the direct shear
strength of this bond on the two samples was determined.
70


Figure 3.7: Deck Core Shear Test Break
The results from the bond shear test are as shown below, in Table 3.3.
Table 3.3: Double Shear Testing Results
Cylinder ID Diameter (in) Area (inA2) SHEAR 1/2 Load Stress (lbs) (psi)
D3A 2.89 6.602 4125 833.0809
D7A 2.89 6.602 4800 969.4032
For the two samples tested, the average shearing stress at the original deck white
topping interface was experimentally determined to be 901.24 psi. Accounting for the
12 samples removed without this bond intact, the consulting engineer determined that
it was safe to assume that 75% of the lower of the two shear strengths could be
transferred between the two layers. This value, 625 psi, is significantly greater than
the 80 psi required in order for composite action to exist (AASHTO 9.20.4.3). Even if
zero shear strength is allotted for each of the 12 samples removed without this bond
in tact, the overall average shear strength for the fourteen samples of 128 psi, still
surpasses the AASHTO requirement. Therefore, a composite section, which included
the concrete topping, could be considered in determining the revised deck rating.
71


3.5 Deck Concrete Core
Ultrasonic Pulse Velocity Testing
Concrete cores removed from the deck were tested using UPV methods to determine
the quality of the concrete. The results of the direct transmission UPV testing through
the length of the core specimens is shown below in Table 3.4. In this table the quality
of concrete is determined by utilizing the correlation with ultrasonic pulse velocity
developed by Ryall, and included as Table 2.1.
Table 3.4: Deck Core Ultrasonic Pulse Velocity Testing Results
Core ID Length (in) Time (M- sec) Length (ft) Velocity (ft / p, sec) Velocity (km / sec) Quality Rating
D6 4.5 38.5 0.37500 0.00974 2.969 Poor
D 2 A 4.34375 28.2 0.36198 0.01284 3.912 Good
This testing was inconclusive in determining the overall condition of the concrete in
the deck, as the results from the limited number of samples varied too greatly to be
utilized in an overall determination of the condition of the original concrete deck.
72


3.6 Deck Testing Conclusions
As previously discussed, the superstructure has insufficient strength rating for the
concrete deck, the transverse steel plate girders, and the longitudinal beam
connections to the transverse steel plate girders. Due to this substandard deck rating
concrete core samples and reinforcing bar samples were removed from the deck for
testing. The findings of this testing were reported to the City and County of Denver
and to the consulting engineer. The following values from the test results were used
to revise the deck rating.
f c = 5,380 psi (the minimum value from the three test samples)
fy = 53,500 psi (the minimum of two non-corroded samples)
The concrete core samples contained approximately 6 inches of the original concrete
deck and an additional 2 inches of the concrete topping. Included in the testing
performed on the deck was the determination of the bond strength between the
concrete topping and the original concrete deck. Based on the test results, it was
determined that a minimum shear stress of 625 psi can be transferred between the two
layers. This is significantly greater than the 80 psi (AASHTO 9.20.4.3) required at a
concrete interface for composite action to exist. Therefore, a composite section,
which included the concrete topping, could be considered in determining the revised
deck rating.
During the removal of the concrete cores from the deck, regions where spacing of the
transverse reinforcing bars did not match the as built drawings. As a result, CCD
contracted a company to complete radiographic examination of the deck.
Approximately three 8-foot sections of the deck were photographed in order to
determine the actual spacing of the transverse reinforcing bars. The average spacing
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of the bars in the photographs was 8 '/g -inches, which is slightly higher than the 8-
inch spacing specified in the existing plans. The spacing of 8 Vg -inches was used to
revise the deck rating.
The revised results indicated that the only substandard inventory rating is the 28.3 ton
rating at the negative moment area over the south exterior girder. It is important to
note that this area of the deck is adjacent to the eastbound bridge and does not have a
median barrier. The median barrier is located on the eastbound bridge, allowing the
truck wheel load to be much closer to the edge of the deck than on the north side of
the bridge. All other locations of the deck have an inventory rating at or near 36 tons.
The City and County of Denver has decided that replacing the deck is not a viable
option at this time. Due to the insufficient rating, the region of the deck over the south
exterior girder must be rehabilitated to extend the life of the deck by 20 to 30 years.
To accomplish this goal, the consulting engineer has proposed two viable options.
The first of these options entails the installation of a series of strips of carbon/epoxy
fiber reinforced polymer (FRP) laminates. The advantage of using this method is that
the repair is localized with minimal impact to traffic.
The other option proposed by the consulting engineer would entail the construction of
a new bridge rail on the south edge of the deck, which would prevent a wheel from
overloading the deck over the south exterior girder. The use of such a barrier would
require a reconfiguration of the lane layout of the bridge.
Most of the testing performed on specimens removed from the deck of the westbound
6th Avenue Viaduct had favorable results. The results from the tensile testing of the
lower level transverse reinforcement samples confirm that the actual tensile strength
is significantly greater than the allowable stress. While the average shear stress at the
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original deck/white topping interface was determined to be considerably greater than
that which is necessary for composite action. The original concrete deck compressive
strength was determined to be greater than the specified ultimate, and finally, the
modulus of elasticity also had favorable results. Based on these results, researchers
were able to provide accurate data to be utilized in modifying the deck rating. Having
done so, a localized area was identified for repairs as it still resulted in a substandard
deck rating after the modifications. This research reassured the City and County of
Denver that they would not have to replace the deck; rather, engineers needed to find
a viable means of increasing the deck rating along the southern edge of the bridge.
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4. Testing of the Pier Columns
According to a preliminary structural analysis, each pier column lacks the flexural
capacity to handle the demand loads. As a result, a consulting engineer has
recommended replacing each pier with a new reinforced concrete pier and pier cap.
However, Piers 13 through 18 are in the UPRR yard and the rehabilitation for these
piers requires approval from UPRR. The UPRR design criterion requires 25 feet of
horizontal clearance from the piers of bridges to the centerline of the tracks. This
requirement applies to all new piers or existing piers that are to be rehabilitated.
Construction is therefore inhibited at Piers 13 and 17 due to the insufficient horizontal
clearance to the railroad tracks. In such cases, where the 25 ft clearance cannot be
provided, a minimum of 9 feet of horizontal clearance is acceptable provided the piers
be protected by a crash wall or be built of heavy construction. If these requirements
cannot be met, a request for variance must be obtained and approved by the UPRR.
Piers 13 and 17 are positioned within the 25 ft UPRR clearance requirements, and
therefore necessitated further testing of insitu elements in order to determine strength
parameters. Testing of these piers was aimed at providing engineering evidence to
decide between leaving the current piers in place, rebuild utilizing either a crash wall
or heavy construction so as to meet the UPRR requirements, or to request variance
from the UPRR for new construction.
Pier 11, the fixed pier, needed to be further examined to determine if the existing pier
could safely hold required loads through the rehabilitation process, to evaluate for
new conditions during and after this rehabilitation, and to verify its need for
replacement.
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It was concluded that new concrete piers and pier caps were to be installed at Piers 1
through 10 and at Piers 12, 14, 15, 16, and 18. At Piers 11, 13 and 17, further
information about the insitu strength and condition of the columns was required to
determine the necessity of repair, and appropriate repair methodology. Therefore,
during the summer of 2004, non-destructive testing was completed on these piers to
determine their strength and condition. Testing consisted of the removal of concrete
core samples from each of the piers in order to perform compressive strength and
modulus of elasticity testing. Four concrete core samples were removed from both
Piers 11 and 13. However, only 3 cores were removed from Pier 17, due to the
bearing geometry of the pier. Additionally, a series of non-destructive testing
techniques were applied to each of the columns in order to develop a comprehensive
understanding as to the insitu condition of each of the columns. This testing consisted
of a visual inspection, hammer sounding, rebound hammer tests, ultrasonic pulse
velocity testing, and tomography.
Prior to testing, a grid was constructed on each of the columns, as shown in Figures
4.1 and 4.2. Starting at the top of the piers, every 2 feet of vertical distance was
marked along the circumference. These lines were labeled such that the 0 level was
located at the top of the pier, the 2 level was 2 feet down from the top of the pier,
labeling continued as such until reaching the ground at the 18 level (18 feet from
the top of the pier). Piers 13 and 17 each have a 5 foot steel pier cap at the top of the
column which precluded testing in this region, and therefore, for these two piers the
0 level was defined as being just below the steel pier cap and continued to the
ground as previously described. The circumference of the column was divided into 8
data points, labeled A-H, such that line A was opposite E, B opposite F, C opposite
G, and D opposite H. Each intersection was then given a unique alpha-numeric
identification (i.e. C-12). This identification grid system was utilized throughout the
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duration of testing to uniquely identify the location of distresses within each of the
columns.
16
18
Figure 4.1: Identification Grid System
Figure 4.2: Identification Grid Layout
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4.1 Pier Columns Visual Inspection
The visual inspection of the 6th Avenue pier columns completed in this study
consisted of carefully examining the surface of the pier columns and mapping the
location of abnormalities, such as cracks or spalls. The most significant observation
made during the visual inspection was a crack surrounding the E-18 data point of Pier
17, as shown in Figure 4.3. However, it is worth restating, that 5-foot steel pier caps
were in place on both Piers 13 and 17, and therefore, no observations could be made
in these regions. Spalling of the concrete at the top of multiple columns prompted this
jacketing effort that placed steel column jackets around the majority of the piers.
Figure 4.3: Pier 17 Crack Encapsulating Data Point El8
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4.2 Pier Columns Hammer Sounding
Hammer sounding testing was completed and recorded for each of the grid points on
each of the columns. In order to graph the results of the qualitative hammer sounding
test, a numeric scale was devised to quantify the results. A good reading was
assigned a value of 1.0, and a bad reading was assigned a value of 0.0. Utilizing
this scale, a contour map of the results was developed using the Surfer software
package.
In order to visualize the results of this surface testing the column was unwrapped to
produce a 2-dimensional image of the column surface as it wraps around the pier
column. The left vertical axis corresponds with the A line continuing alphabetically
to the right where the right vertical axis corresponds with the H line. Hammer
Sounding Testing results are presented in Figures 4.4 to 4.6.
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Figure 4.4: Pier 11 Hammer Sounding Testing Results
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ABCDEFGH
1 l 1 J I 5
Figure 4.5: Pier 13 Hammer Sounding Testing Results
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1
0.9
0.8
0.7
0.6
0.5
04
0.3
0.2
0.1
0
Figure 4.6: Pier 17 Hammer Sounding Testing Results
The limited locations that tested bad for the hammer sounding testing are clearly
identifiable in Figures 4.4 to 4.6. The poor testing results recorded at point E-18 of
Pier 17 came as no surprise after having observed the crack encompassing this data
point. However, there were no such visual signs of distress around data points A-0
and E-0 of Pier 11, the only other locations to test negatively. Inevitably, further
testing would reveal the depth of the damage at these locations.
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4.3 Pier Columns Rebound Hammer Testing
For the purposes of the 6th Avenue westbound study, an average of five rebound
numbers was recorded for each grid location on each of the piers, as shown in Figure
6. The recorded rebound number was the average value displayed on the unit.
The schmidt hammer results are presented as a contour graph of the average rebound
numbers at each of data points for each of the pier columns. Since the rebound
number is proportional to the compressive strength of the concrete, no conversion
was necessary in order to produce a graph of the relative strength. A high rebound
number corresponds with greater strength and a lower rebound number with lesser
strength. The results are presented as being relative to one another such that areas of
relative strength and weakness can be identified in each of the pier columns.
Similar to the hammer sounding results, the schmidt hammer results are presented in
a 2-dimensional image of the column surface as it wraps around the pier column.
Schmidt hammer results are presented in Figures 4.7 to 4.9.
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As shown in Figure 4.7, Schmidt hammer testing of Pier 11 indicated lower average
rebound numbers, and thus decreased strength at the following locations:
D-F 0-6, 16
A 1-7, 11-15
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As shown in Figure 4.8, Schmidt hammer testing of Pier 13 indicated lower average
rebound numbers, and thus decreased strength at the following locations:
C-D 2-6
F-H 2-6
B&D 10-14
F-H 14
F18
86